1
BEHAVIOR OF CONCRETE BRIDGE DECKS REINFORCED
2
WITH HIGH-PERFORMANCE STEEL
3
Hatem M. Seliem, Gregory Lucier, Sami H. Rizkalla, and Paul Zia
4 5 6 7 8 9 10 11 12 13
Hatem M. Seliem (corresponding author),
[email protected] Gregory Lucier,
[email protected] Sami H. Rizkalla,
[email protected] Paul Zia,
[email protected] North Carolina State University Constructed Facilities Laboratory (CFL) 2414 Campus Shore Dr. Raleigh, NC 27695-7533, USA
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Authors’ Biographies
15 16 17 18 19 20 21 22 23 24 25 26 27 28 29 30 31 32 33 34
ACI Member Hatem M. Seliem received his B.Sc. and M.Sc. degrees, with honors, from Cairo University, Egypt in 2000 and 2002, respectively. Currently, he is a Ph.D. candidate in the Department of Civil, Construction, and Environmental Engineering at North Carolina State University. ACI Member Gregory Lucier received his B.Sc. and M.Sc. degrees, with honors, from North Carolina State University in 2004 and 2006, respectively. Currently, he is working as a research engineer at the Constructed Facilities Laboratory, North Carolina State University. ACI Fellow Sami H. Rizkalla is Distinguished Professor of Civil and Construction Engineering in the Department of Civil, Construction and Environmental Engineering, North Carolina State University. He serves as the Director of the Constructed Facilities Laboratory and NSF I/UCRC in Repair of Structures and Bridges at North Carolina State University. He is a fellow of ACI, ASCE, CSCE, EIC and IIFC. An ACI honorary member and past president, Paul Zia is Distinguished University Professor Emeritus at North Carolina State University. He served as ACI president in 1989, and is a member of several ACI Committees including ACI 363, High-Strength Concrete; joint ACIASCE 423, Prestressed Concrete; ACI 445, Shear and Torsion; the Concrete Research Council; TAC-TTC; and serves as chairman of TTC-ITG6.
1
1
Abstract
2
This paper describes the behavior of concrete bridge decks reinforced with newly developed
3
High-Performance (HP) steel which is characterized by its high strength and enhanced corrosion-
4
resistance in comparison to conventional ASTM A 615-061 Grade 60 steel. The selected HP steel
5
is commercially known as Micro-Composite Multi-Structural Formable (MMFX) steel. The
6
study presented herein included testing of three full-scale bridge decks with a span-to-depth ratio
7
of 12.5. The first and second decks were constructed with the same reinforcement ratio using HP
8
and Grade 60 steel, respectively. The third deck was reinforced with HP steel using 33 percent
9
less reinforcement in an attempt to utilize its high strength. A non-linear finite element model
10
was used to predict the mode of failure and failure loads. Test results demonstrate that the use of
11
HP steel at a reduced reinforcement ratio is viable as flexural reinforcement in concrete bridge
12
decks. The paper also presents the test results of specially-designed specimens to study the effect
13
of bending of HP steel bars on their tensile strength.
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Keywords
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high-performance; high-strength; bridge decks; bent bars; punching; flexural-shear
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INTRODUCTION
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Bridge decks are frequently subjected to severe environmental conditions which often lead to
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serious corrosion problems. The use of High-Performance (HP) steel could help to mitigate
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corrosion problems due to its enhanced corrosion resistance. In addition, HP steel has higher
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strength compared to conventional ASTM A 615-061 Grade 60 steel. Therefore, by using HP
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steel the amount of required reinforcement could be considerably reduced. Reducing the amount 2
1
of steel will alleviate reinforcement congestion and improve concrete placement. Commercially
2
available steel known as Micro-Composite Multi-Structural Formable (MMFX) steel, which
3
conforms to ASTM A 1035-072, was selected for this study because of its high-strength and
4
enhanced corrosion resistance in comparison to conventional ASTM A 615-061 Grade 60 steel.
5
This paper is a part of a comprehensive study to investigate the structural behavior of HP steel
6
for bridges. The work presented in this paper examined the behavior of bridge deck slabs and the
7
strength of bent bars required for certain details. The experimental program presented in this
8
paper consisted of two phases. In the first phase, three full-scale bridge decks with a span-to-
9
depth ratio of 12.5 were tested to evaluate the structural performance of bridge decks reinforced
10
with HP steel as main flexural reinforcement in comparison to the use of conventional Grade 60
11
steel. In the second phase, four specially-designed specimens were tested to assess the effect of
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bending on the tensile strength of HP steel bars.
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RESEARCH SIGNIFICANCE
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Recently, many state transportation departments have begun to use HP steel as a direct
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replacement for conventional Grade 60 steel in concrete bridge decks3. However, the behavior of
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concrete bridge decks reinforced with this novel steel is not well defined. This study is an
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attempt to utilize the high strength characteristics of HP steel in concrete bridge decks. In
18
addition, the study evaluates the effect of bending on the tensile strength of HP steel bars.
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PHASE I: CONCRETE BRIDGE DECKS
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Test Specimens
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A total of three full-scale bridge decks were considered in this study to examine the flexural limit
22
state behavior including the mode of failure. The three decks were designed to be identical in all 3
1
aspects except for the type and amount of steel used in each. All three bridge decks consisted of
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two spans and double cantilevers, supported in composite action by three precast, post-tensioned
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concrete girders having cross-sectional dimensions of 24x10 in. (610 x 254 mm). The overall
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nominal dimensions of the bridge decks were 21’-10”x13’-2”x 8⅝” (6655 x 4013 x 220 mm)
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with a span-to-depth ratio of 12.5. The supporting girders were post-tensioned using deformed
6
prestressing bars of 1 in. (25 mm) diameter with an ultimate strength of 150 ksi (1034 MPa).
7
Each girder was prestressed by four bars resulting in a total prestressing force of 360 kips (1601
8
kN) per girder. Post-tensioning was used to prevent the girders from torsional cracking so as to
9
maintain their torsional stiffness throughout the test. The girders were designed so that their
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torsional stiffness was similar to that of the steel bridge girders of an actual bridge that was built
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in Johnston County, North Carolina in 2004 using HP steel3.
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The first and third bridge decks were reinforced with HP steel, while the second bridge deck was
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reinforced with conventional Grade 60 steel for comparison purposes. The test matrix is given in
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Table 1, and the reinforcement details for the three bridge decks are shown in Fig. 1. It should be
15
noted that the reinforcement ratio (ρ) is calculated using the total slab thickness. The first and
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second bridge decks were constructed with the same reinforcement ratio using HP and
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conventional Grade 60 steel similar to that used in the bridge built in Johnston County, North
18
Carolina, 20043. However, the third bridge deck was reinforced with HP steel using only two-
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thirds of the reinforcement ratio used for the first two decks. The reduction of the amount of steel
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is based on a selected yield strength of 90 ksi (621 MPa) which is within the linear behavior of
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the HP steel and less than the yield strength of 120 ksi (827 MPa) determined according to the
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0.2 % offset method specified by ASTM A 370-074. It should be noted that only the transverse
23
steel was reduced since the deck is continuous in this direction where primary bending occurs. 4
1
Material Properties
2
The three bridge decks were constructed using normal-weight concrete with average
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compressive strengths at the day of testing for the three bridge decks of 7000, 4500, and 5300 psi
4
(48, 31, and 36 MPa), respectively. The concrete compressive strength was determined using 4 x
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8 in. (102 x 204 mm) concrete cylinders cast for each deck and cured under the same conditions
6
as the deck. Tension coupons of HP and Grade 60 steels were tested according to ASTM A 370-
7
074. The stress-strain characteristics of the HP and Grade 60 steel are shown in Fig. 2. The HP
8
steel reinforcing bars exhibit a linear stress-strain relationship up to 100 ksi (689 MPa) followed
9
by a nonlinear behavior up to an ultimate strength of 173 ksi (1193 MPa). According to the
10
ASTM A370-074 offset method (0.2 % offset), the yield strength was determined to be 120 ksi
11
(827 MPa). The initial modulus of elasticity was 29,000 ksi (200 GPa), followed by a nonlinear
12
behavior and reduction in the modulus of elasticity when the stress exceeded 100 ksi (689 MPa).
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The yield strength of the Grade 60 steel was determined to be 68 ksi (469 MPa).
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Test Setup and Instrumentation
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Two 440 kips (1957 kN) capacity MTS hydraulic actuators were used to apply a concentrated
16
load to each span simultaneously to simulate the effect of truck wheel loads. Two 10 x 20 in.
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(254 x 508 mm) steel plates of 1 in. (25 mm) thickness were used to transfer the loads from the
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actuators to comply with the AASHTO Specifications5 for tire contact area. A ½ in. (13 mm)
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thick neoprene pad was placed under each loading plate to prevent possible local crushing of the
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concrete. The supporting girders rested on concrete blocks to transfer the applied loads to the
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strong floor resulting in a clear span of 96 in. (2438 mm). The clear span of supporting girders
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was determined based on the equivalency of the torsional stiffness of the supporting girders to
5
1
that of the steel girders used in the actual bridge. Fig. 3 shows an isometric view of the test setup
2
and a photograph of the first bridge deck prior to testing.
3
A total of 72 channels of instrumentation were used on each bridge deck. A 550 kips (2447 kN)
4
capacity load cell was mounted to each actuator to measure the applied load. Twenty-four string
5
potentiometers (string pots) were used to measure the deflection profiles of the bridge deck along
6
the longitudinal and transverse directions. In addition, six linear potentiometers were used to
7
measure the deflections and rotations at the mid-span of each girder. Twenty PI gages (also
8
known as wire arch strain gauges, see Fig. 4(a)) were used to measure the concrete strain at
9
various locations. Twenty electrical resistance strain gages of 120 ohm and 6 mm gage length
10
were attached to selected reinforcing bars to determine the strains in these bars. Data were
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electronically recorded by an Optim Megadac data acquisition system. Fig. 4 shows the locations
12
of the PI gages used and establishes the notation used hereafter.
13
Test Results
14
Test results were analyzed to examine critically the performance of bridge decks reinforced with
15
HP steel as main reinforcement compared to the behavior of bridge decks reinforced with Grade
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60 steel. Detailed test results can be found elsewhere3.
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Load-Deflection Behavior
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The three bridge decks were subjected to loading and unloading to load levels of 50, 100, and
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150 kips (222, 445, and 667 kN) per span, and then to failure. The load-deflection envelopes up
20
to failure for the three tested bridge decks are shown in Fig. 5. It should be noted that the
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deflections shown in Fig. 5 are measured at the center of the respective deck span directly under
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the applied load. It can be seen from Fig. 5 that the first bridge deck reinforced with HP steel 6
1
using the same reinforcement ratio as used for the actual bridge exhibited smaller deflections
2
than that of the other two bridge decks at the same load level. The slightly higher stiffness of the
3
first deck is likely due to the higher concrete compressive strength, and to the higher strength of
4
HP steel. Despite the lower reinforcement ratio used for the third bridge deck (33 percent less
5
than that of the first two decks), it was capable of achieving the same ultimate load-carrying
6
capacity as the second bridge deck reinforced with the Grade 60 steel. This behavior is attributed
7
to the higher tensile strength of HP steel. The slight increase of the deflection measured for the
8
third bridge deck compared to the second deck is due to the use of less steel and to the slight
9
reduction of the modulus of elasticity of HP steel at high stress levels.
10
The deflection profiles along the transverse direction for the right span of the second and third
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bridge decks are shown in Fig. 6. It should be noted that the deflection profiles are plotted for the
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last loading cycle only. Therefore, residual deflections are shown at the beginning of the loading
13
cycle (zero load). The deflection profiles indicate that the maximum deflection occurred at mid-
14
span under the applied load. The deflection profiles also show that the deflection behavior of the
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deck reinforced with reduced amount of HP steel is very similar to that of the deck reinforced
16
with conventional Grade 60 steel.
17
Deflection profiles in the longitudinal direction for the right span of the second and third bridge
18
decks are given in Fig. 7. It should be noted that the deflection profiles are plotted for the final
19
loading cycle only. Accordingly, the deflections shown in Fig. 7 include the residual deflections
20
from previous loading cycles. The longitudinal deflection profiles demonstrate the curvature in
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the longitudinal direction, implying the two-way flexural behavior of typical bridge decks under
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concentrated loads. In addition, the deflection profiles illustrate that the deflection at the edge of
7
1
the decks was very small. This indicates that selection of the length of the test models is adequate
2
for carrying the total load, and therefore, representative to typical bridge decks.
3
Crack Pattern
4
No cracks were observed up to a load level of 50 kips (222 kN) for any of the three bridge decks.
5
The first visible top cracks occurred at a load level of approximately 60 kips (267 kN) for each
6
deck. According to the AASHTO Specifications5, an axle of the design truck consists of a pair of
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16 kip (71 kN) wheel loads spaced 6’ (183 mm) apart. Therefore, at a load level of 21 kips (93
8
kN), which includes the dynamic allowance, the three bridge decks remained un-cracked and the
9
deflection at service load level was identical for the three bridge decks. Therefore, reducing the
10
amount of HP steel used in the third bridge deck did not alter the serviceability behavior.
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As expected, negative flexural cracks formed before positive flexural cracks due to the higher
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induced negative moments at the middle support in comparison to the positive moment values at
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mid-span. Positive moment flexural cracks were observed at load levels of 100 kips (445 kN)
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and became radial at a load of 150 kips (667 kN), as shown for the first bridge deck in Fig. 8 (a)
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and Fig. (b), respectively. Formation of the radial crack pattern confirms the two-way
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mechanism under the effect of the concentrated applied loads. Further loading led to spreading
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and widening of the flexural cracks until the formation of flexural-shear cracks at the top surface
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of the deck close to the middle girder. The formation of the flexural-shear cracks led to a sudden
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drop in the load as shown in Fig. 5. However, the flexural-shear cracks formed symmetrically on
20
both sides of the middle girder of the first bridge deck and allowed an increase of the load and
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finally failure of the deck by punching shear in both spans. For the second and third bridge
8
1
decks, the flexural-shear crack occurred only on the left side of the middle girder. This allowed
2
the load to increase until failure occurred due to punching shear in one of the spans only.
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Mode of Failure
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Due to the selection of sufficient length of the test model, the behavior of the bridge decks under
5
the effect of the concentrated loads was two-way flexural mechanism followed by development
6
of an arching action supported by membrane forces developed in the bottom layer of the
7
reinforcement. Due to the continuity used in the test models, at the measured first peak of load
8
for the first bridge deck a sudden drop in the load resistance occurred due to the formation of the
9
flexural-shear cracks along the top surface of the bridge deck on both sides of the middle girder.
10
Further loading caused widening of those cracks associated with slight increase in the load
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resistance until punching failure occurred under the applied concentrated loads. Punching failure
12
of both spans occurred almost simultaneously at load levels of 229 kips (1019 kN) and 216 kips
13
(961 kN), and corresponding deflections of 1.8 in. (46 mm) and 1.6 in. (41 mm) for the left and
14
right spans, respectively. Fig. 9 shows the first bridge deck at the conclusion of the test, where
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the punching areas under the loads and the shear cone at the bottom of the left span can be seen.
16
The behavior of the second bridge deck, reinforced with Grade 60 steel using the same
17
reinforcement ratio was similar to the first deck. At the measured peak load, a sudden drop in the
18
applied load occurred due to the formation of the flexural-shear crack on the top surface of the
19
bridge deck to the left side of the middle girder only. Failure of the left span was due to flexure-
20
shear failure which prevented an increase in load sufficient to cause punching shear. The
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maximum measured load for the left span was 185 kips (823 kN) and maximum deflection was
9
1
2.2 in. (56 mm). Failure of the right span was due to punching shear at a load level of 204 kips
2
(907 kN) and corresponding deflection of 0.7 in. (18mm).
3
For the third bridge deck, the right span failed in punching shear at a load of 203 kips (903 kN)
4
and a corresponding deflection of 1.0 in. (25 mm) prior to the failure of the left span. Formation
5
of the flexural-shear crack in the left span caused a sudden drop in the applied load and
6
prevented the capability to increase the applied load to induce punching shear failure in the left
7
span. Failure of the left span was due to flexural-shear failure at a load level of 181 kips (805
8
kN) and the maximum measured deflection was 1.9 in. (48 mm). Fig. 10 shows the second and
9
third bridge decks at failure, where the punching area under the applied concentrated load at mid-
10
span and the flexural-shear crack formed within the vicinity of the middle girder are clearly
11
visible.
12
Strain in Concrete and Steel
13
Based on the deformations measured by the PI gages, concrete compressive strains were
14
determined. Concrete strain is plotted for the final loading cycle only, and hence, includes the
15
residual strain developed in previous loading cycles. The strain obtained from the PI gage
16
located in the right span of the three bridge decks 14 in. (356 mm) from the centerline of the
17
deck (T6 in Fig. 4(a)) are shown in Fig. 11. The measured strain from the second deck indicates
18
that the compressive strain of the concrete in the vicinity of the punching area reached the
19
limiting value of 0.002 typically observed by others to cause punching shear failure.6,7,8
20
However, the first and third bridge deck compressive strains exceeded this value reaching values
21
of 0.0031 and 0.0036, respectively. In order to explain this behavior, the strain in the reinforcing
22
steel has to be investigated. 10
1
The steel strains were measured using conventional electrical strain gages attached to the
2
reinforcing bars before casting. The strains measured in the bottom transverse steel bars of the
3
right span of the first and second bridge decks are shown in Fig. 12. The steel strains are shown
4
for the final loading cycle only, and therefore, include residual strains developed in previous
5
loading cycles. It can be seen from Fig. 12(a) that the steel bar at mid-span of the first bridge
6
deck did not reach the yielding strain of HP steel bars, as defined earlier. On the other hand, Fig.
7
12(b) shows that the strain in the bar located at mid-span of the second bridge deck exceeded the
8
yielding strain of Grade 60 steel. Collected data from the various strain measurements indicate
9
that yielding was very localized in the vicinity of the concentrated load3.
10
It is well established that the concrete in the vicinity of the shear cone is under a triaxial state of
11
stress.6,7,8 For the second bridge deck, the steel in the vicinity of the punching cone had yielded
12
prior to failure. Therefore, the steel bars were no longer restraining the concrete, and failure
13
occurred when the concrete reached its peak stress at a corresponding strain of 0.002. For the
14
first and third bridge decks, the transverse and longitudinal compressive stresses were
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continuously increasing up to failure since the steel in the vicinity of the punching cone did not
16
yield. Therefore, the steel bars were still restraining the concrete and the concrete was still intact,
17
thus maintaining the aggregate interlock across the shear crack. Consequently, the measured
18
concrete strain was still increasing until the concrete crushed in the vicinity of the shear crack at
19
a strain level much higher than 0.002, reaching values of 0.0031 and 0.0036 for the first and third
20
bridge decks, respectively.
11
1
Predicted Punching Capacity
2
The predicted punching shear strengths for the three bridge decks according to two different
3
design codes are summarized along with the measured values in Table 2. The design codes
4
presented in Table 2 are the AASHTO Specifications5 and the American Concrete Institute (ACI
5
318-05).9 The design equations used for the predictions are:
6
⎡ ⎤ 0.126 ;0.126⎥ f c' b ο d AASHTO: Vc = min ⎢0.063 + βc ⎣ ⎦
units: kips & in.
Equation (1)
7
⎡ ⎤ αd 4 ACI: Vc = min ⎢2 + ;4; s + 2⎥ f c' b ο d bο ⎣ βc ⎦
units: lbs & in.
Equation (2)
8
where Vc = punching shear capacity of bridge deck; βc = ratio of long side to short side of
9
loading plate; fc’ = concrete compressive strength; bо = perimeter of critical section at a distance
10
of d/2 from loading plate; d = effective section depth; and αs = constant.
11
It is clear from Table 2 that the predicted values according to the AASHTO and ACI 318-05
12
design codes compare very well to the measured values for bridge decks reinforced with HP and
13
Grade 60 steel.
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ANALYTICAL MODELLING
15
The three bridge decks tested in this study were modeled analytically using the finite element
16
analysis program “ANACAP” (Anatech Concrete Analysis Program) Version 3.0.10 The concrete
17
material model is based on smeared cracking model11. As stated in the program manual, this
18
model is a mechanics-based formulation using plasticity theory that permits incorporation of
19
cracking and other concrete response characteritstics.10 Within the concrete constitutive model,
20
cracking and all other forms of material non-linearity are treated at the finite element integration 12
1
points. Cracks are assumed to form perpendicular to the principal tensile strain direction in
2
which the criterion is exceeded and they are allowed to form at each material point. When
3
cracking occurs, the normal stress across the crack is reduced to zero and distribution of stresses
4
around the crack is recalculated. Concrete modeling also included residual tension stiffness for
5
the gradual transfer of load to the reinforcement during crack formation. In addition, the program
6
accounts for the reduction in shear stiffness due to cracking and further decay as the crack opens.
7
The reinforcement is modeled as individual sub-elements within the concrete elements. The
8
stiffness of the bar sub-element is superimposed on the concrete element stiffness in which the
9
bar resides. The anchorage loss is modeled as an effective stiffness degradation of the bar as a
10
function of the concrete strain normal to the bar.
11
A 3-D analysis was conducted for the three bridge decks using 20-node hexahedral continuum
12
elements with quadratic isoparametric displacement interpolation. Only one quarter of the deck
13
was modeled due to its symmetry about both axes.
14
including mesh size (number and size of elements) and loading increment. The depth of the deck
15
was divided into five layers within its thickness with a total number of elements of 1040 for the
16
deck and supporting girders, as shown in Fig. 13(a).
17
Analytical Results
18
The predicted and experimental load-deflection envelopes for the three bridge decks are
19
compared in Fig. 13. It can be seen that the predicted load-deflection behaviors of the three
20
bridge decks compared very well with the measured values. The initial and post-cracking
21
stiffnesses were accurately predicted by the analytical model. In addition, the ultimate load was
22
also reasonably predicted considering the fact that the two spans of the second and third bridge 13
A convergence study was conducted
1
decks failed in two different modes. However, the predicted ultimate deflections were slightly
2
less than the experimental values, which is due to the fact that the program was terminated when
3
the concrete strain in the compression zone reached the value of 0.003. For validation purposes,
4
the portion of the first bridge deck that failed due to punching shear was cut using a concrete saw
5
to reveal the failure mode as shown in Fig. 14 (a). The strain contours shown in Fig. 14 (b)
6
depicting the punching shear cone, match very well to the actual failure mode.
7
PHASE II: TENSILE STRENGTH OF HIGH-PERFORMANCE STEEL BENT BARS
8
Test Specimens and Test Setup
9
A specially-designed specimen was used to evaluate the effect of bending HP steel bars on their
10
tensile strength; the specimen used is shown in Fig. 15(a) for #5 (No. 16) bars. The specimen
11
consisted of two concrete blocks used to anchor the two ends of the bent bar in the shape of a
12
closed stirrup. The two sizes of steel reinforcement used for this phase were #4 (No. 13) and #5
13
(No. 16) bars with two specimens for each size. The bend was 90 degrees according to ACI 318-
14
05,9 and the lengths of the HP steel stirrups were selected based on the dimensions of the
15
concrete blocks, the dimensions of the hydraulic jack, and the load cell placed between the
16
concrete blocks. The concrete blocks were heavily reinforced with conventional Grade 60
17
stirrups to prevent premature failure. The blocks were cast using wooden forms which were
18
specially designed to accommodate the anchored ends and temporarily braced to prevent bending
19
of the exposed portion of the bar between the two blocks before testing.
20
Two different configurations were used to debond the steel from the concrete within the same
21
specimen. In the first configuration, the stirrup was completely debonded in the left concrete
22
block by using a thick rubber tape. In the second configuration, only the straight portion of the 14
1
stirrup was debonded in the right concrete block to transfer the tension force directly to the bent
2
portion of the bar. This study is a continuation to a previous study12 that was conducted in 2002
3
at the North Carolina State University using the same specimen. In the previous study, only the
4
straight portions of the stirrups were debonded rather than debonding the entire U-shaped section
5
of the bar for the specimens reported in this paper. Debonding of the entire U-shaped portion
6
allows relative movement of the bar with respect to the concrete. This movement allowed pure
7
testing of the bent bars rather than representing typically bonded bars and stirrups used for
8
concrete structures. The results obtained from the previous study are compared to those obtained
9
from the work reported here to demonstrate the effect of bond to the concrete on the tensile
10
strength of bent bars.
11
The test setup, shown in Fig. 15(b), consists of a 120 kips (534 kN) capacity hydraulic jack, a
12
150 kips (667 kN) capacity load cell, and four linear potentiometers to measure the relative
13
displacement between the two blocks. The hydraulic jack and the load cell were centered
14
between the two branches of the stirrup to ensure equal distribution of forces in each branch. An
15
MTS axial mechanical extensometer of 2 in. (51 mm) gage length was mounted on the exposed
16
portion of the stirrup to measure the elongation during loading. An OPTIM Megadac data
17
acquisition system was used to electronically record the readings of the load cell, the
18
potentiometers, and the extensometer.
19
Test Results
20
Failure of all the specimens for both bar sizes occurred at the bent portion of the stirrup which
21
was debonded from the concrete in the left block, as shown in Fig. 16(a). This failure location
22
was confirmed by visual inspection after sawing the concrete blocks at the location of failure as 15
1
shown in Fig. 16(a). It should be noted that failure of the specimens tested in the previous study12
2
(with bonded bends) occurred at the exposed straight portion of the stirrup as shown in Fig.
3
16(b). Typical stress-strain characteristics of #4 (No.13) and #5 (No.16) debonded bent bars
4
along with those of bonded bars, are shown in Fig. 17. The stress-strain characteristics of the
5
bonded and debonded bent bars indicate that their behavior is similar to the bonded bars,
6
including the linear and the non-linear behavior up to a strain value of 1.5 percent. However,
7
testing of the debonded bent bars emphasizes the induced residual strains due to bending of the
8
bars which reduced both the strength and the strain at ultimate. This behavior reflects the well
9
established phenomenon of the stress concentration at the bend location due to the bending
10
process. It should be mentioned that in actual structures, bent bars are always bonded to the
11
concrete enabling them to reach the ultimate stress and strain of straight bars as reported
12
previously.12 Based on these tests, the results suggest that HP steel bars can be bent up to 90
13
degrees without affecting their ultimate strength or strain, provided that the bend is fully encased
14
and bonded to concrete.
15
CONCLUSIONS AND RECOMMENDATIONS
16
In light of the test results, the following conclusions can be drawn:
17
1. The ultimate load-carrying capacity of the three bridge decks investigated in this study was
18
on the order of ten times the service load prescribed by the AASHTO Specifications.5
19
2. Punching shear was the primary mode of failure for the three bridge decks. Due to continuity
20
used in the test models, flexural-shear failure was observed as a secondary mode of failure.
21
3. The cracking load of the three tested bridge decks was more than twice the service load
22
prescribed by the AASHTO Specifications.5 Hence, under service load level, the three bridge 16
1
decks behaved as un-cracked sections. Therefore, using 33 percent less HP steel should not
2
alter the serviceability behavior of concrete bridge decks.
3
4. The bridge deck reinforced with 33 percent less HP steel developed the same ultimate load-
4
carrying capacity as that reinforced with Grade 60 steel. This performance is attributed to the
5
higher strength of HP steel compared to Grade 60 steel.
6
5. Behavior of bonded HP steel bent bars is similar to the behavior of straight bars. Debonded
7
bent bars exhibit similar behavior to straight bars, including the linear and the non-linear
8
behavior up to a strain of 1.5 percent. However, its ultimate strength is reduced by 6 percent
9
and its ultimate strain by 70 percent.
10
Based on the research findings the following design guidelines can be recommended:
11
1. Substituting HP steel directly for conventional Grade 60 steel in a design, as was done for the
12
actual bridge built in Johnston County, North Carolina, in 2004 and the first test specimen is
13
a conservative approach.
14
2. HP steel can be used as the main flexural reinforcement for cast-in-place concrete bridge
15
decks at a reinforcement ratio corresponding to 33 percent less than that required for Grade
16
60 steel. Therefore, design of reinforced concrete bridge decks using HP steel can utilize a
17
yield stress of 90 ksi (621 MPa) for the HP steel bars.
18
3. Reduced reinforcement ratio of HP steel shall satisfy all minimum reinforcement ratios
19
prescribed by the AASHTO Specifications.5 In addition, the reduced reinforcement ratio of
20
HP steel must comply with the crack control requirement of the AASHTO Specifications.5
21
4. HP steel bars can be bent up to 90 degrees without reducing their ultimate strength or strain
22
provided that the bend is fully encased and bonded to concrete.
17
1
ACKNOWLEDGMENT
2
The authors gratefully acknowledge the support of the North Carolina Department of
3
Transportation (NCDOT) for sponsoring this research project. Thanks are due to MMFX
4
Technologies Corporation for supplying the MMFX steel, to CC Mangum Inc. of Raleigh, NC
5
for helping to cast the three bridge decks, and to SteelFab of Charlotte, NC for donating the steel
6
sections used in the testing frames. Special thanks are extended to Catrina Walter for taking part
7
in the second phase of the experimental program. The authors would also like to thank Jerry
8
Atkinson and Bill Dunleavy at the Constructed Facilities Laboratory for their untiring help.
9
REFERENCES
10
1. ASTM A 615/A 615M-06, “Standard Specification for Deformed and Plain Carbon-Steel for
11
Concrete Reinforcement,” ASTM International, West Conshohocken, PA, 2006, 6 pp.
12
2. ASTM A 1035/A 1035M-07, “Standard Specification for Deformed and Plain, Low-Carbon,
13
Chromium, Steel Bars for Concrete Reinforcement,” ASTM International, West
14
Conshohocken, PA, 2007, 5 pp.
15
3. Rizkalla, S. H., Zia, P., Seliem, H. M., and Lucier, G., “Evaluation of MMFX Steel for
16
NCDOT Concrete Bridges,” Research Report, North Carolina State University, Constructed
17
Facilities Laboratory, Raleigh, NC,USA, December 2005, 109 pp.
18 19 20 21
4. ASTM A 370-07, “Standard Test Methods and Definitions for Mechanical Testing of Steel Products,” ASTM International, West Conshohocken, PA, 2003, 47 pp. 5. American Association of State Highway and Transportation Officials, “AASHTO LRFD Bridge Design Specifications”, Washington, D.C., Third Edition, 2004.
18
1 2 3 4 5 6
6. Kinnunen, S., and Nylander, H., “Punching of Concrete Slabs without Shear Reinforcement,” Transactions of the Royal Institute of Stockholm, Sweden, 1960, 112 pp. 7. Marzouk, H. and Hussein, A., “Punching Shear Analysis of Reinforced High-Strength Concrete Slabs,” Canadian Journal of Civil Engineering, V. 18, 1991, pp. 954-963. 8. Mufti, A and Newhook, J., “Punching Shear Strength of Restrained Concrete Bridge Deck Slabs,” ACI Structural Journal, V. 95, No. 4, July-August, 1998, pp. 375-381.
7
9. ACI Committee 318 (ACI 318-05), “Building Code Requirements for Structural Concrete
8
(ACI 318-05) and Commentary (ACI 318R-05),” American Concrete Institute, Farmington
9
Hills, MI, 2005, 430 pp.
10 11 12 13
10. James, R.G., “ANACAP Concrete Analysis Program Theory Manual, Version 3.0, Anatech Corporation, San Diego, CA, 2004. 11. Rashid, Y.R., “Ultimate Strength Analysis of Prestressed Concrete Pressure Vessels”, Nuclear Engineering and Design, 1968, pp. 334-344.
14
12. El-Hacha, R. and Rizkalla, S. H., “Fundamental Material Properties of MMFX Steel Bars,”
15
Research Report 02-04, North Carolina State University, Constructed Facilities Laboratory,
16
Raleigh, NC, July 2002, 60 pp.
17 18 19 20 21 22 23 19
1
Table 1: Bridge decks test matrix Bridge Deck
Steel Type
First
Second
Third
Bottom Reinforcement
Top Reinforcement
Transverse
Longitudinal
Transverse
Longitudinal
HP
#5 @ 6.75 in. No. 16 @ 170 mm ρ = 0.54%
#5 @ 10 in. No. 16 @ 250 mm ρ = 0.36%
#5 @ 6.75 in. No. 16 @ 170 mm ρ = 0.54%
#4 @ 14 in. No. 13 @ 360 mm ρ = 0.17%
Grade 60
#5 @ 6.75 in. No. 16 @ 170 mm ρ = 0.54%
# 5 @ 10 in. No. 16 @ 250 mm ρ = 0.36%
#5 @ 6.75 in. No. 16 @ 170 mm ρ = 0.54%
#4 @ 14 in. No. 13 @ 360 mm ρ = 0.17%
HP
#5 @ 10 in. No. 16 @ 250 mm ρ = 0.36%
#5 @ 10 in. No. 16 @ 250 mm ρ = 0.36%
#5 @ 10 in. No. 16 @ 250 mm ρ = 0.36%
#4 @ 14 in. No. 13 @ 360 mm ρ = 0.17%
2 3 4
Table 2: Predicted and experimental punching shear strength Bridge Deck
Steel Type
Transverse Reinforcement Ratio %
Experimental
AASHTO
ACI
First
HP
0.54
229 (1019)
229 (1019)
230 (1023)
Second
Grade 60
0.54
204 (907)
184 (818)
184 (818)
Third
HP
0.36
203 (903)
199 (885)
200 (890)
Punching Shear Strength, kips (kN)
5 6
20
#4 @ 14" (#13 @ 360)
85 8" [220 mm]
#5 @ 6¾" (#16 @ 170)
#5 @ 10" (#16 @ 250)
#25 Deformed Prestressing Bars
2' [610 mm]
#5 @ 6¾" (#16 @ 170)
#25 Deformed Prestressing Bars
#25 Deformed Prestressing Bars
2' [610 mm]
4 Branches #4 (#13) #25 Deformed Prestressing Bars
2' [610 mm]
4 Branches #4 (#13)
4 Branches #4 (#13) #25 Deformed Prestressing Bars
#25 Deformed Prestressing Bars
10" [254 mm]
10" [254 mm]
10" [254 mm]
9' [2743 mm]
1'-11" [584 mm]
9' [2743 mm]
1'-11" [584 mm]
21'-10" [6655 mm]
Reinforcement Details of the First and Second Bridge Decks #4 @ 14" (#13 @ 360)
85 8" [220 mm]
#5 @ 10" (#16 @ 250)
#5 @ 10" (#16 @ 250)
#25 Deformed Prestressing Bars
2' [610 mm]
#25 Deformed Prestressing Bars
2' [610 mm]
4 Branches #4 (#13) #25 Deformed Prestressing Bars
#25 Deformed Prestressing Bars
2' [610 mm]
4 Branches #4 (#13)
4 Branches #4 (#13)
#25 Deformed Prestressing Bars
10" [254 mm] 1'-11" [584 mm]
#5 @ 10" (#16 @ 250)
#25 Deformed Prestressing Bars
10" [254 mm] 9' [2743 mm]
10" [254 mm] 9' [2743 mm]
1'-11" [584 mm]
21'-10" [6655 mm]
1
Reinforcement Details of the Third Bridge Deck
2
Fig. 1: Reinforcement details for the three bridge decks 180
1,200 High-Performance
150
Stress (ksi)
120
800 Grade 60
90
600
60
400
30
200
0 0
3 4
0.03
0.06
0.09
0.12
0 0.15
Strain (in/in)
Fig. 2: Stress-strain characteristics of HP and Grade 60 steel bars 21
Stress (MPa)
1,000
(a) Test setup
1
(b) First bridge deck prior to testing
Fig. 3: Test setup and the first bridge deck prior to testing
2
(a) Locations of PI gages
3
(b) Schematic plan
Fig. 4: Locations of PI gages and notation for the three bridge decks
4
22
Mid-Span Vertical Deflection (mm) 13 25 38 51
0
Mid-Span Vertical Deflection (mm) 13 25 38 51
0
64
64
250
250
1000
1000
100
400
50
First Bridge Deck-MMFX Second Bridge Deck-G60 Third Bridge Deck-MMFX
AASHTO LRFD Service Load
0 0
0.5 1 1.5 2 Mid-Span Vertical Deflection (in.)
Span Load (kips)
600
Span Load (KN)
Span Load (kips)
800 150
800 150 600 100
400
50
200 0
0
2.5
0
0.5 1 1.5 2 Mid-Span Vertical Deflection (in.)
(a) Left span
1
200
First Bridge Deck-MMFX Second Bridge Deck-G60 Third Bridge Deck-MMFX
AASHTO LRFD Service Load
Span Load (KN)
200
200
0
2.5
(b) Right span
Fig. 5: Load-deflection envelopes of the three bridge decks
2 15"
9"
10"
10"
9"
15"
20"
C
23"
15"
9"
10"
10"
9"
15"
20"
23"
1
25
1
25
0.5
13
0.5
13
-0.5
-13
-1
-25
Span Load = 0 kips Span Load = 50 kips (222 KN) Span Load = 100 kips (445 KN) Span Load = 150 kips (667 KN) Span Load = 200 kips (889 KN) Span Load = 204 kips (907 KN)
-2
-2.5 0
-38 -51
Vertical Deflection (in.)
0
0
-1.5
0
0
-0.5
-13
-1
Span Load = 0 kips Span Load = 50 kips (222 KN) Span Load = 100 kips (445 KN) Span Load = 150 kips (667 KN) Span Load = 200 kips (889 KN) Span Load = 203 kips (903 KN)
-1.5 -2
-2.5
-64 15 30 45 60 75 90 105 120 135 Distance from Longitudinal CL of Bridge Deck (in.)
0
15
30
45
60
75
90
105
120
-25 -38 -51
-64 135
Distance from Longitudinal CL of Bridge Deck (in.)
(a) Second bridge deck
3
20"
Vertical Deflection (mm)
20"
Vertical Deflection (mm)
Vertical Deflection (in.)
C
(c) Third bridge deck
Fig. 6: Transverse deflection profiles for the right spans of the second and third bridge decks
4
23
32.5"
20
CL
32.5"
40
60
80
Vertical Deflection (in.)
0 -0.5
-13
-1
-25
-1.5 -2 -2.5
1
0 0
Span Load = 0 kips Span Load =50 kips (222 KN) Span Load = 100 kips (445 KN) Span Load = 150 kips (667 KN) Span Load = 200 kips (889 KN) Span Load = 204 kips (907 KN)
-38 -51
5"
9"
32.5"
20
32.5"
40
60
0
-0.5
-13
-1
-25 Span Load = 0 kips
-1.5
Span Load =50 kips (222 KN)
-38
Span Load = 100 kips (445 KN)
-2
Span Load = 150 kips (667 KN)
-51
Span Load = 200 kips (889 KN) Span Load = 203 kips (903 KN)
-64
-2.5
-64
Distance From Transverse CL of Bridge Deck (in.)
Distance From Transverse CL of Bridge Deck (in.)
(a) Second bridge deck
(c) Third bridge deck
Fig. 7: Longitudinal deflection profiles for the right spans of the second and third bridge deck
2 3
(a) At load level of 100 kips (445 kN)
4
80
0
Vertical Deflection (mm)
9"
Vertical Deflection (in.)
0
5"
Vertical Deflection (mm)
CL
(b) At load level of 150 kips (667 kN)
Fig. 8: Positive flexural cracks for the first bridge deck
5
24
(a) First bridge deck at the conclusion of the test
1
(b) Punching cone in the left span
Fig. 9: Failure of the first bridge deck
2
(a) Second bridge deck
3
(b) Third bridge deck
Fig. 10: Failure of the second and third bridge decks
25
Strain (in/in) -0.004
-0.003
-0.002
-0.001
0
Frist Deck MMFX
1000 Second Deck Grade 60
Span Load (kips)
200
800
Third Deck Reduced MMFX
150
600 ε
100
400 First Bridge Deck_MMFX
50
Span Load (kN)
-0.005 250
200
Second Bridge Deck_Grade 60 Third Bridge Deck-Reduced MMFX 0 -5000
0 -4000
-3000 -2000 Strain (Micro-Strain)
1 2
-1000
0
Fig. 11: Concrete compressive strain in vicinity of the punched area for the three bridge decks
3 Steel Strain (in/in) 0
0.001
0.002
0.003
0.004
0.005
0
0.006
Steel Strain (in/in) 0.002 0.003 0.004
0.005
0.006
250
250
Edge
Quarter Span
Mid-Span
1000
Edge
1000
Quarter Span Mid-Span
600
100
400 Mid-Span Quarter Span Edge
50 0 0
1000
2000 3000 4000 Steel Strain (Micro-Strain)
5000
800
Span Load (kips)
150
Span Load (KN)
Span Load (kips)
800
150 600
100
400 Mid-Span
50
200
Quarter Span
Span Load (KN)
200
200
200
Edge
0 6000
0 0
(a) First bridge deck
4
0.001
1000
2000 3000 4000 Steel Strain (Micro-Strain)
5000
0 6000
(b) Second bridge deck
Fig. 12: Strain in bottom transverse steel for the right spans of the first and second bridge decks
5
26
0
Mid-Span Vertical Deflection (mm) 12.7 25.4 38.1 50.8
63.5
250 1000 800 150 600 100
400
50
Left Span Right Span ANACAP
0 0
(a) Mesh used for the analytical model
0
Mid-Span Vertical Deflection (mm) 13 25 38 51
0.5 1 1.5 2 Mid-Span Vertical Deflection (in.)
64
0
0
2.5
Mid-Span Vertical Deflection (mm) 13 25 38 51
64
250 1000
1000
600 100
400
50
Left Span Right Span ANACAP
0 0
0.5 1 1.5 2 Mid-Span Vertical Deflection (in.)
800 150 600 100
400
50
200 0
Left Span Right Span ANACAP
0
2.5
0
(c) Second bridge deck
0.4 0.8 1.2 1.6 2 Mid-Span Vertical Deflection (in.)
200 0
2.4
(d) Third bridge deck
Fig. 13: Mesh, analytical and experimental load-deflection envelopes for the three bridge decks
2
27
Span Load (KN)
800 150
Span Load (kip)
200 Span Load (KN)
200 Span Load (kip)
200
(b) First bridge deck
250
1
Span Load (KN)
Span Load (kip)
200
(a) Experimental
1
(b) ANACAP
Fig, 14: Principal strain contours at failure for the first bridge deck
2 3
(a) Schematic plan view of the bent bars test setup
4
(b) Test setup
Fig. 15: Schematic details and test setup for bent bar specimens
28
(a) Debonded
1
(b) Bonded
Fig. 16: Location of failure of #4 (No. 13) debonded and bonded bent bar specimens
2 3
Bonded
150
Extensometer Removed
1,200
180
1,000
150
1,200
120
800
90
600
60
400
30
200
0 0
4
0.01
0.02
0.03
0.04
0.05
Stress (ksi)
Bonded
Stress (MPa)
Stress (ksi)
Debonded
0 0.06
120
Extensometer Removed
Debonded
800
90
600
60
400
30
200
0 0
0.01
0.02
0.03
0.04
Strain (in/in)
Strain (in/in)
(a) #4 (No. 13) bars
(b) #5 (No. 16) bars
0.05
Fig. 17: Stress-strain relationship for debonded and bonded bent bars
5
29
1,000
0 0.06
Stress (MPa)
180
