Behavior of Double Skin Composite Wall Subjected to

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Tae-Sung Eom1; Hong-Gun Park2; Cheol-Ho Lee3; Jin-Ho Kim4; and In-Hwa Chang5 Abstract: Double skin composite walls are composed of two steel plate “skins” connected by tie bars, with the space between them filled with concrete. They were developed to reduce wall thickness, to enhance constructability, and to enable rapid construction by eliminating the use of formwork and reinforcing bars. In the present study, cyclic testing was performed to investigate the seismic behavior of isolated and coupled double skin composite walls with rectangular and T-shaped cross sections. The wall specimens failed mainly by tensile fracture of the welded joints at the wall base and coupling beams, or by local buckling of the steel plates. Because of their large depth, the ductility of the wall specimens was not as good as that of beams having less depth. In particular, the ductility of the walls was significantly affected by the strengthening methods used for the wall base. The load-carrying capacities of the isolated and coupled wall specimens were evaluated using plastic stress distributions in their cross sections, which provided satisfactory predictions. DOI: 10.1061/共ASCE兲ST.1943-541X.0000057 CE Database subject headings: Cyclic loads; Seismic tests; Coupled walls; Shear walls; Composite structures.

Introduction Structural walls have long been used as an efficient lateral forceresisting system for buildings. Although, traditionally, walls have been constructed with reinforced concrete, steel plate walls are a promising alternative. Steel plate walls can reduce the weight of the structure, increase the usable floor area by the use of relatively thin walls, and enable rapid construction. Three types of steel plate walls have been studied: framed steel plate wall 关Fig. 1共a兲, Driver et al. 共1998兲, Park et al. 共2007兲兴, profiled composite framed wall 关Fig. 1共b兲, Hossain and Wright 共2004a,b,c, 2005兲兴, and double skin composite wall 关Fig. 1共c兲, Wright et al. 共1991a,b兲兴. The framed steel plate wall comprises a steel moment frame and an infill steel plate. The profiled composite framed wall comprises a boundary steel frame, profiled steel sheets connected to the steel frame, and infill concrete. Experimental tests were performed for the framed steel plate wall and 1

Lecturer, Dept. of Architecture, Catholic Univ. of Daegu, 330, Kuemnak-ri, Hayang-eup, Gyeongsan-si, Gyeongbuk, South Korea 712702. E-mail: [email protected] 2 Associate Professor, Dept. of Architecture, Seoul National Univ., San 56-1, Shinlim-Dong, Kwanak-Gu, Seoul, South Korea 151-744. E-mail: [email protected] 3 Professor, Dept. of Architecture, Seoul National Univ., San 56-1, Shinlim-Dong, Kwanak-Gu, Seoul, South Korea 151-744. E-mail: [email protected] 4 Head Researcher, Research Institute of Industrial Science and Technology, 79-5, Yeongcheon-ri, Dongtan-myeon, Hwasung-si, Kyeonggido, South Korea. E-mail: [email protected] 5 Head Researcher, Research Institute of Industrial Science and Technology, 79-5, Yeongcheon-ri, Dongtan-myeon, Hwasung-si, Kyeonggido, South Korea. E-mail: [email protected] Note. This manuscript was submitted on August 5, 2008; approved on April 6, 2009; published online on September 15, 2009. Discussion period open until March 1, 2010; separate discussions must be submitted for individual papers. This paper is part of the Journal of Structural Engineering, Vol. 135, No. 10, October 1, 2009. ©ASCE, ISSN 0733-9445/ 2009/10-1239–1249/$25.00.

the profiled composite framed wall. Based on the results, analysis and design methods have been developed. A double skin composite wall comprises two steel plates connected by tie bars 共Wright et al. 1991a,b兲 and an infill of concrete. The double skin composite wall is a good choice when thin and uniform wall thickness is required. The composite action of the concrete and steel plates provides strength and stiffness for the wall. The filled concrete prevents early buckling of the steel plates, while the steel plates are used as the formwork for the concrete filler during construction. Wright et al. 共1991a兲 tested double skin composite beams and columns subjected to static loading. Based on their test results, they developed design methods for double skin composite members, which are similar to the design methods used for doubly reinforced concrete members 共Wright et al. 1991b兲. The local buckling of the steel plate skins and the shear strength of the tie bars connecting the steel plate skins have been evaluated by Wright et al. 共1991b兲, Wright 共1995兲, and Liang et al. 共2004兲. Xie and Chapman 共2006兲 developed detailed design methods for double skin composite beams subjected to static loading and fatigue loading. These previous studies focused on double skin composite beams and columns. In the present study, we investigated the structural capacity of double skin composite walls as a lateral force-resisting system. Cyclic tests were performed for both slender isolated walls and coupled walls. The seismic resistance of the walls including the load-carrying capacity and ductility was evaluated.

Experimental Program Three isolated walls and two coupled walls subjected to cyclic lateral loading were tested. The test specimens are shown in Figs. 2共a–e兲. Isolated walls and coupled walls were 1/3 and 1/4 scale models of the core wall for a 30-story prototype building, showing flexure-dominated behavior, respectively. The configurations and section properties of the specimens were determined based on the results of elastic analysis and structural design for the proto-

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Fig. 1. Steel plate walls

type building with the double skin composite walls. Though the prototype core wall was subjected to both lateral and gravity loads, axial compressive load was not considered in this test. DSCW1 共DSCW1N, DSCWH, and DSCWC兲 was tested three times with different strengthening methods used at the wall base. Fig. 2共f兲 shows DSCW1N, which was fabricated with tie bars 共bolted connections兲, including a view inside the wall. As presented in Table 1, test parameters were the wall type, cross-sectional shape, steel plate thickness, and strengthening method used for the wall base. For the isolated wall specimens DSCW1 共DSCW1N, DSCW1H, DSCW1C兲, DSCW2, and DSCW3, 10-mm-thick SN490 共Korean Standard designation兲 steel plates were used. The yield strength and ultimate strength of the SN490 steel plates, Fy and Fu, obtained from tensile coupon tests, were 383 and 544 MPa, respectively. For the coupled wall specimens DSCW4 and DSCW5, 5.9-mm-thick SPAR295 共Korean Standard designation兲 steel plates were used. The yield strength and ultimate strength of the SPAR 295 steel plates were measured to be 372 and 453 MPa, respectively. The compressive strengths f ⬘c of the filled concrete were 39.7 and 58.7 MPa for the isolated and coupled wall specimens, respectively. Electrode with a specified minimum Charpy V-Notch toughness of 26.75 J at −28.9 ° C was specified for the flux-cored arc welding. Fig. 2共a兲 shows the details of the isolated wall specimen DSCW1 共DSCW1N, DSCW1H, and DSCW1兲 which had a rectangular cross section. The steel walls were fabricated by connecting steel plates with complete joint penetration groove welds. In the test for DSCW1N, which was not strengthened at the wall base, premature brittle failure occurred at the wall base due to tensile fracture of the welded joint before the flexural yielding of the wall. To study the effects of various strengthening methods for the wall base on cyclic seismic behavior, after the test for DSCW1N, the specimen was strengthened with different methods, and the tests were re-executed until a satisfactory strengthening method was finally found. For DSCW1H, triangular rib plates 共SN490, thickness= 10 mm, height= 150 mm兲 were welded to both ends of the cross section at the wall base 关Fig. 2共a兲兴. For DSCW1C, the wall thickness at the wall base was uniformly increased by using enclosing cover plates 共SN490, thickness= 10 mm, height= 280 mm兲 关Fig. 2共a兲兴. These cover plates were fillet-welded to both the wall plates and the base plate. Specimen DSCW2 was given a vertical joint at the center of the cross section along the wall height 关Fig. 2共b兲兴. This vertical

joint was bolt-connected with connecting steel plates 共160⫻ 60 ⫻ 10 mm兲 at spacings of 300 mm. To strengthen the vertical joint, horizontal shear bars 共␸ 16 mm兲 were embedded in the filled concrete, at the same spacings. Complete joint penetration groove welding was used at the vertical joint to within 700 mm of the wall bottom. Specimen DSCW3 had a T-shaped cross section. At the vertical joint between the web wall and flange wall, fillet welding was used and shear bars 共␸ 16 mm兲 were embedded in the filled concrete at spacings of 300 mm 关Fig. 2共c兲兴. Both DSCW2 and DSCW3 were strengthened by using cover plates 共SN490, thickness= 10 mm, height= 280 mm兲 at the wall base. Figs. 2共d and e兲 show coupled wall specimens DSCW4 and DSCW5, which had rectangular and T-shaped cross sections, respectively. The coupling beams were connected to the wall plates by full penetration groove welding. During the test for DSCW4, early brittle fracture occurred at the welded connections between the coupling beams and the wall plates. Therefore, in DSCW5, the joints between the coupling beams and the wall plates were strengthened with cover plates 关100⫻ 300⫻ 5.9 mm, Fig. 2共e兲兴. The concrete in the wall and coupling beams was cast monolithically. DSCW4 and DSCW5 were also strengthened by cover plates 共SPAR295, thickness= 5.9 mm, height= 130 mm兲 at the wall base. For DSCW5, which had a T-shaped cross section, the vertical joint between the web wall and flange wall was connected in the same manner as used in DSCW3. Wright 共1995兲 reported that in double skin composite beams subjected to monotonic loading, when the ratio of the tie bar spacing to steel plate thickness was less than 37, there was no brittle failure due to local buckling of the steel plates. In the present study, tie bars connecting two steel plate skins were arranged at vertical and horizontal spacings of 300 mm, which were about 30 共isolated wall specimens兲 and 50 共coupled wall specimens兲 times the plate thickness. At the wall base and at the joints between the wall plate and the coupling beams, where significant inelastic deformation was expected to develop, the nuts for the tie bars were welded to the steel plates in order to prevent the fracture of the steel plates, which might have been triggered by a bolt slipping or loosening. For lateral loading, the top of the specimens was strengthened by welding three horizontal steel plates 共10 mm thickness兲 to the wall 关see B-B section in Fig. 2共a兲兴. However, in the coupled walls DSCW4 and DSCW5, the horizontal steel plates were not welded to the top coupling beam, to prevent the horizontal steel plates from contributing to the coupling action 关see B-B section in Figs.

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Fig. 2. Double skin composite wall specimens 共length unit: mm兲

2共d and e兲兴. Therefore, the top coupling beam was designed to have the same shear-carrying capacity as those of the other coupling beams. A 3,000-kN actuator was used to apply cyclic lateral loads to

the top of the wall specimens. The loadings were controlled by the horizontal displacement at the top of the walls. Fig. 3 shows the loading history for the tests. From elastic analysis, the yield displacement ⌬y at the top of the DSCW1 was estimated to be 38

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Table 1. Test Parameters

DSCW1N DSCW1H DSCW1C DSCW2 DSCW3 DSCW4 DSCW5

Plate thickness 共steel type兲

Wall type

10 mm 共SN490兲

Isolated wall

5.9 mm 共SPAR295兲

Coupled wall

Shape of cross section

Dimension of cross section 共width⫻ depth兲

Rectangle Rectangle Rectangle Rectangle T-shape Rectangle T-shape

120 120 120 120

mm⫻ 1 , 000 mm⫻ 1 , 000 mm⫻ 1 , 000 mm⫻ 1 , 000 See Fig. 2共c兲 100 mm⫻ 1 , 500 See Fig. 2共e兲

mm 共1.0% drift ratio兲. Based on the predicted yield displacement ⌬y 共38 mm兲, the target displacement was incremented by ⫾ 0.5 ⌬y 共19 mm or 0.5% drift ratio兲: ⫾ 0.5, 1.0, 1.5, 2.0, 2.5, 3.0, and 3.5%. Two load cycles were repeated at each target displacement. The loading speed was 0.2 mm/sec. The tests were terminated when the load-carrying capacity of the specimens decreased to 80% of the maximum capacity.

Experimental Results The lateral load versus top displacement relationships and the envelope curves of the specimens are shown in Fig. 4. The yield displacement ⌬y, maximum strength Vu, maximum displacement ⌬u, ductility ␮, and failure modes of the specimens were summarized in Table 2. For the calculation of the ductility, the yield displacement was defined by using the concept of equal plastic energy 关Fig. 4共g兲, Choi and Park 共2008兲兴 so that the area enclosed by the idealized elastic-perfectly plastic envelope curve was the same as that by the actual envelope curve. The maximum displacement was defined as the postpeak displacement at the instant when the strength was degraded to 80% of the peak strength. Fig. 5 shows the failure mechanisms of the specimens. DSCW1N, which was not strengthened at the wall base, failed due to brittle tensile fracture of the welded joint, at +1.5% drift ratio before flexural yielding of the wall 关Figs. 4共a兲 and 5共a兲兴. Because of the early fracture at the wall base, DSCW1N showed slip deformation in the load-displacement relationship 关Fig. 4共a兲兴. DSCW1H, which was strengthened with triangular rib plates, also failed due to fracture of the welded connection at +1.5% drift ratio, though by using the rib plates the maximum strength was increased and slip deformation did not occur 关Fig. 4共b兲 and 5共b兲兴. On the other hand, DSCW1C, which was strengthened with cover plates, showed a stable cyclic behavior up to +2.52% drift ratio, without fracture of the welded joints 关Fig. 4共c兲兴. At ⫺2.52% drift ratio, DSCW1C failed due to outward buckling of the steel plates

 

 

 

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Specimen

 









 









     

Fig. 3. Loading history

Ratio of tie bar spacing to plate thickness

Strengthening method for wall base

mm mm mm mm

30

mm

50

None Rib plate Cover plate Cover plate Cover plate Cover plate Cover plate

and subsequent fracture of the vertical weld joint, followed by crushing of the filled concrete and fracture of the tie bars 关Fig. 5共c兲兴. DSCW2, which had a vertical joint at the center of the wall, showed stable cyclic behavior up to ⫺2.0% drift ratio. However, severe twisting of the wall occurred during the negative loading. This test result appears to imply that a stronger connecting scheme is required for this type of vertical joint. However, in actual buildings, the severe twisting of walls is not expected to occur because of the diaphragm rigidity of the slabs. After the twisting of the wall, monotonic loading was applied only in the positive direction 关Fig. 4共d兲兴. Ultimately, under this monotonic loading, DSCW2 failed at +7.1% drift ratio by the combined failure modes of wall buckling, local buckling of the steel plate, and tie bar fracture 关Fig. 5共d兲兴. This result demonstrates that the ductility of a double skin composite wall under cyclic loading is significantly less than that of a comparable wall under monotonic loading. In DSCW3, which had a T-shaped cross section, severe strength degradation occurred when the web wall failed from local buckling of the steel plate, crushing of the filled concrete, and tie bar fracture, at ⫺2.0% drift ratio 关Fig. 5共e兲兴. After the strength degradation, monotonic loading was applied in the positive direction. At +4.9% drift ratio, DSCW3 failed by tensile fracture of the steel plate just above the wall base which had been strengthened by cover plates 关Fig. 5共e兲兴. Like DSCW2, the maximum displacement of DSCW3 under monotonic loading was 2.5 times the maximum displacement under cyclic loading. In DSCW4, the coupled wall specimen with two rectangular cross sections, strength degradation occurred at +1.5% drift ratio, from buckling of the steel plate and crushing of the filled concrete. At +2.0% drift ratio, the vertical welded joint was fractured by the repeated bending of the buckled steel plate 关Fig. 5共f兲兴. Tensile cracking at the welded connection between the wall plate and the coupling beam was initiated at +1.5% drift ratio. DSCW4 共s / t = 50兲 which had a greater ratio of tie bar spacing 共s兲 to steel plate thickness 共t兲, failed earlier than DSCW1C 共s / t = 30兲. As shown in Fig. 4共g兲, DSCW5, the coupled wall composed of two T-shaped walls, showed stable cyclic behavior up to +2.5% drift ratio. At +3.0% drift ratio, its load-carrying capacity was significantly degraded due to local buckling of the steel plate and fracture of the welded connections in the coupling beams. Because of the cover plate strengthening used 关Fig. 2共e兲兴, tensile cracking at the welded joints for the coupling beams did not occur until 2.0% drift ratio. Buckling of the steel plate at the bottom of the wall was initiated at +1.5% drift ratio, but did not proceed further. Fig. 6 shows the lateral load versus steel strain relationships of the specimens. The strains in the steel plates were measured with strain gages, which were installed vertically at the ends of the

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Fig. 4. Lateral load versus top displacement relationships of specimens

cross section, at 300 mm 共DSCW1C and DSCW3兲 or 150 mm 共DSCW4 and DSCW5兲 away from the wall bottom. In DSCW1C, large compressive plastic strains developed in the steel plates at both ends of the cross section 关Fig. 6共a兲兴, although they were less

than the maximum tensile strains. This result indicates that the filled concrete was subjected to a substantial compressive force. In DSCW3, the isolated wall specimen with the T-shaped cross section, the neutral axis of the cross section was located close to

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Table 2. Summary of Test Results

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Specimens

Max. strength Max. displ. Yield displ. Max. drift Ductility ratio ⌬u共mm兲 ⌬y共mm兲 ⌬u / l共%兲 a Loading type ⌬u / ⌬ y Vu共kN兲

DSCW1N 667 58 DSCW1H 765 59 DSCW1C 869 97 DSCW2 共+兲 809 273 DSCW2 共⫺兲 796 78 1,291 189 DSCW3 共+兲 b 1,212 77 DSCW3 共⫺兲c DSCW4 822 76 DSCW5 1,390 100 a l = 3850 mm. b 共+兲: flange wall in compression. c 共⫺兲: flange wall in tension. d Final monotonic loading after cyclic loading.

— — 34 — 35 — 44 21 27

1.51 1.53 2.52 7.08 2.02 4.91 2.00 1.97 2.59

the flange wall. Thus, the tensile strain at the end of the web wall was greater than that in the flange wall 关Fig. 6共b兲兴. In DSCW4, the coupled wall specimen with the thin steel plates thickness = 5.9 mm, local buckling of the steel plates developed early at +1.5% drift ratio, and the buckled steel plates experienced repeated bending during subsequent cyclic loading. As a result, the strain curve of the steel plates in Fig. 6共c兲 was significantly distorted. In DSCW5, the coupled wall specimen with the T-shaped

— 2.86 — 2.21 — 1.78 — 3.63 3.69

Cyclic Cyclic Cyclic C-Md Cyclic C-Md Cyclic Cyclic Cyclic

Failure modes Fracture at welded connection at wall base Fracture at welded connection at wall base Plate buckling, concrete crushing, tie bar fracture Wall buckling, plate buckling, tie bar fracture Twisting of top beam Tensile fracture of steel plate Plate buckling, concrete crushing, tie bar fracture Plate buckling, concrete crushing Fracture of plates in coupling beam

cross section, the maximum tensile strain of the wall was less than those of other specimens, because the neutral axis was located close to the flange wall 关Fig. 6共d兲兴. Further, unlike DSCW1C, DSCW3, and DSCW4, the strains in the steel plates of DSCW5 remained in tension during the entire cyclic loading because the filled concrete of the flange wall was able to resist the large compressive force.

Fig. 5. Failure modes of double skin composite wall specimens 1244 / JOURNAL OF STRUCTURAL ENGINEERING © ASCE / OCTOBER 2009

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Fig. 6. Lateral load versus steel strain relationships

Considerations for Ductility The results of this testing program showed that double skin composite walls with a large depth are susceptible to tensile fracture at the welded joints in the regions of high stress concentration, along with local buckling of the steel plates. In particular, the specimens that were not well strengthened at the wall base showed unsatisfactory ductile behavior. Therefore, one of the most important considerations when designing double skin composite walls that will be subjected to severe seismic loading should be assuring their ductility. Wall Configuration and Effective Depth The ductility of the walls tested in this study was not as good as that of shallow beams. Fig. 7 shows the variation in the plastic rotational capacity of the welded steel moment connections, including those of DSCW1C, DSCW3, DSCW4, and DSCW5, according to their depth. As shown in the figure, the plastic

rotational capacities of the isolated and coupled walls were measured using LVDTs installed at the plastic hinge region. Generally, the plastic rotational capacity of welded steel moment connections consistently decreases as the beam depth increases 共FEMA 2000a; Roeder 2002兲. One of the primary reason for this is that, for a given curvature or rotation, a greater strain demand develops in the extreme fibers of the cross section as the member depth increases. This observation implies that to increase the ductility of the wall through minimizing brittle fracture, the effective wall depth 共the distance from the neutral axis to the extreme fiber of the wall section兲, which determines the magnitude of the maximum strain and stress, should be decreased if possible. For example, in DSCW5, the coupled wall composed of two T-shaped walls, the maximum strain of the wall section was relatively small, because the neutral axis was located close to the flange wall 关see Fig. 6共d兲兴. On the other hand, DSCW3, the isolated T-shaped wall, exhibited less ductility because of its larger effective depth.

Fig. 7. Variation of plastic rotation of steel members according to member depth JOURNAL OF STRUCTURAL ENGINEERING © ASCE / OCTOBER 2009 / 1245

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Strengthening Method at Wall Base Since the 1994 Northridge earthquake, the possibility of premature brittle fracture of unreinforced welded steel moment connections has been recognized. Because of this, the current seismic design criteria for steel buildings are based on the capacity design concept 共Bruneau et al. 1998; FEMA 200b; AISC 2005兲. One of the strategies that have been proposed to protect the more vulnerable welded joints, for the capacity design concept, is to make the connection stronger than the member framing used for the connection. Cover plates, ribs, side plates, and haunches have been successfully implemented to strengthen the connection 共Engelhardt and Sabol 1998; Lee et al. 2005; Uang et al. 2000; Lee et al. 2003兲. As described previously, two strengthening schemes were evaluated in this study. DSCW1N, which was not strengthened at the wall base, failed due to early fracture of the welded joint and developed very poor ductility. The test result for DSCW1H 关Fig. 4共a兲兴 showed that the rib plate strengthening method 关Fig. 2共a兲兴 was not effective in preventing early fracture of the welded joints. This is because using the rib plates, or a light and local strengthening scheme, was not fully effective in reducing stress concentration, by increasing the flexural capacity of the wall base. On the other hand, DSCW1C, DSCW3, DSCW4, and DSCW5, which were strengthened with cover plates, exhibited much better performances and failed at 2.0–3.77% drift ratio 共ductility ⌬u / ⌬y = 1.78– 3.69, Table 2兲. These test results imply that a rather heavy and redundant strengthening scheme which does not increase stress concentration should be used for the wall base. The test result for DSCW5 showed that the cover plate strengthening used to connect the coupling beams to the wall also worked well. Of course, in a seismic design, the strengthening should be done in conjunction with the use of notch tough weld filler metal. As discussed, in this study, the cover plate strengthening method was recommended to make the wall base stronger than the wall. However, for full scale walls having 3–4 times the wall depth tested, fracture of the welded joint is expected to become severe because the plastic strains become larger. In this case, it is recommended to extend the double skin composite wall to the basement or the foundation under the ground level. If so, the flexural moment at the wall base can be significantly decreased by the lateral supports of the basement slabs and grade beams, and thus, premature failure at the welded joint at the wall base can be prevented. Local Buckling of Steel Plates As shown in the test results for DSCW1C, DSCW3 共negative loading兲, and DSCW4, when early fracture of the welded connections was prevented, the double skin composite walls subjected to cyclic loading primarily failed from local buckling of the steel plates and subsequent concrete crushing. Fig. 8 shows the mechanism for the steel plate buckling under repeated tension compression. In the steel plates subjected to reversed cyclic loading, a large compressive stress developed early, even in the tensile strain, due to the residual tensile strain. However, since the filled concrete could not resist compression in the tensile strain, the steel plate became susceptible to buckling. Once buckling of the steel plates occurred, excessive compressive force was imposed on the filled concrete. Further, the buckled steel plates could not provide lateral confinement to the filled concrete. Thus, the crushing of the concrete and fracture of the tie bars followed the buckling of the steel plates. For this reason, the double skin composite

Fig. 8. Buckling mechanism of steel plate under cyclic loading

walls subjected to cyclic loading failed earlier than the walls subjected to monotonic loading. As shown in Figs. 6共a and c兲, the buckling of the steel plates and subsequent concrete crushing developed in DSCW1C, DSCW3, and DSCW4, and as a result, large compressive strains developed in the steel plates. Fig. 6 shows the tensile strains in the steel plates at the time the steel plate buckling was initiated and at the time the wall specimens failed due to the plate buckling. In DSCW1C and DSCW3, which had 10-mm-thick steel plates 共tie bar spacing-tosteel plate thickness ratio= 30兲, steel plate buckling occurred at tensile strains of 0.021–0.028 mm/mm. On the other hand, in DSCW5, which had 5.9-mm-thick steel plates 共tie bar spacing-tosteel plate thickness ratio= 50兲, steel plate bucking occurred at smaller tensile strains, 0.014–0.015 mm/mm 共The test result for DSCW4, which had the same steel plate thickness, was less reliable because the strain records were distorted by the local buckling of the steel plates兲. The crushing of the filled concrete and the fracture of the tie bars following the steel plate buckling caused failure of the walls. DSCW1C and DSCW3 共tie bar spacing-tosteel plate thickness ratio= 30兲 failed at tensile strains of 0.041– 0.043 mm/mm. As shown in Figs. 4共d and e兲, in the walls subjected to monotonic loading 共DSCW2 and 3 under positive loading兲, early buckling of the steel plates did not occur.

Load-Carrying Capacity If the early buckling of the steel plates was restrained, and tie bars were used with sufficient shear strength to prevent excessive bond slip between the steel plates and the filled concrete, the loadcarrying capacity of a slender double skin composite wall could be evaluated by using the plastic stress distribution at the cross section 共Wright et al. 1991b; Xie and Chapman 2006兲. Fig. 9 shows the plastic stress distribution at the wall cross sections. The yield strength f y and the effective strength f ce 共=0.85f ⬘c 兲 were used for the plastic stresses of the steel plate and concrete, respectively. Equations for the calculation of the plastic moment capacity M n are presented in Table 3. In Table 3 and Fig. 9, c is the depth of the compression zone, which is calculated from the forceequilibrium in the cross section; b and h are the thickness and depth of the wall cross section, respectively; t is the plate thickness; b f and h f are the width and thickness of the flange wall 共T-shaped cross section兲; xc is the distance from the centroid to the flange end 共T-shaped cross section兲; ␤ 共=0.85兲 is the factor defining the depth of the equivalent concrete stress block; Pe is the axial compressive force applied to the walls 共positive and negative values for compression and tension, respectively兲; and n 共=Es / Ec兲 is the ratio of the elastic modulus Es of steel to the elastic modulus Ec of concrete.

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Fig. 10. Force transfer mechanism in coupled wall

Fig. 9. Plastic stress distribution at double skin composite cross sections

For isolated walls 共DSCW1C, DSCW2, and DSCW3兲 showing cantilever action, the plastic moment of the cross section could be calculated using the plastic strengths of the steel plates and concrete. Then, the load-carrying capacity of the wall could be calculated by dividing the plastic moment by the shear span of the wall. The overall load-carrying capacity of the coupled walls 共DSCW4 and DSCW5兲 was defined by the sum of the plastic moments of the two separate walls and the contribution of the coupling axial forces 共Fig. 10兲 Vn =

Nlw + M Tn + M Cn ⌺Vbnlw + M Tn + M Cn = ls ls

共1兲

where Vn = overall load-carrying capacity of the coupled wall; Vbn = shear transfer capacity of each coupling beam; and ls = shear span of the wall. The plastic moments of the two walls, M Tn and M Cn, can be calculated with equations in Table 3 considering the axial tension and compression force N 共=Pe兲 developed by the coupling beam. The shear transfer capacity Vbn of the coupling beam can be calculated as follows. During cyclic loading, longitudinal elongation develops in the coupling beams due to the tensile residual strain of the steel reinforcement 共Paulay 1996兲. This elongation causes tensile crack-

ing in the concrete. As a result, the concrete with tensile cracks cannot provide adequate resistance to the shear force applied to the beam, though it can prevent the buckling of the steel plates. Thus, the majority of the shear force is resisted by the steel plates 共Lam et al. 2005兲. In the present study, we calculated the shearcarrying capacity of the double skin composite coupling beam, ignoring the contribution of the filled concrete. The shear-carrying capacity of a double skin composite coupling beam is determined as the minimum of the capacities by flexural yielding and shear yielding 关British Steel Institute 共BSI兲 1990; Lam et al. 2005兴 Vbn = min共Vbf ,Vbs兲

共2兲

where Vbf = maximum shear capacity developed by flexural yielding 共=2M bp / lb兲; M bp = plastic moment of the coupling beam 共=ZFy兲; Z = plastic section modulus; lb = length of the coupling beam 共Fig. 10兲; Vbs = maximum capacity by shear yielding 共=AveFy / 冑3兲; Ave = effective sectional area for shear resistance 共=2hbtb兲; hb = depth of the coupling beam; and tb = thickness of the steel plate used in the coupling beam. Table 4 presents the predicted load-carrying capacities, Vn, of the wall specimens. For these predictions, the distance from the critical section to the top beam was used as the shear span ls. Here, the critical sections were located at 280 mm from the wall base for the isolated walls and at 130 mm for the coupled walls. For the coupled walls 共DSCW4 and DSCW5兲, the shear-carrying capacity Vbn of the coupling beam was determined as the capacity Vbs 关=507 kN⬍ Vbf 共=543 kN兲兴 by shear yielding. It was as-

Table 3. Plastic Moment Capacity of Double Skin Composite Sections Depth of compression zone

Cross-sectional shape Rectangular

— Compression in flange wall 共positive loading兲

Plastic moment capacity M n = bhtf y + 2c共h − c兲tf y + 共 b / 2 − t兲␤c共h − ␤c兲f ce

c ⱕ hf

M n = 关共h − c兲2 + b f h f + bh + c2兴tf y + 共xc − c兲Pe + b f c2␤共1 − ␤ / 2 兲 f ce

c ⬎ hf

M n = 关共h − c兲2 + b f 共2c − h f 兲 + b共h − c兲 + c2兴tf y + 共xc − c兲Pe + 关共h f − 2t兲共b f − b兲共c − ht / 2 兲 + 共b − 2t兲c2␤共1 − ␤ / 2 兲兴 f ce

T-shaped Tension in flange wall 共negative loading兲

c ⱕ h − hf

M n = 关共h − c兲2 + b f 共2h − 2c − h f 兲 + bc + c2兴tf y + 共h − c − xc兲Pe + 共b − 2t兲c2␤共1 − ␤ / 2 兲 f ce

c ⬎ h − hf M n = 关共h − c兲2 + b f h f + bc + c2兴tf y + 共h − c − xc兲Pe + 共b − 2t兲c2␤共1 − ␤ / 2 兲 f ce 2 Note: x = 关共b f h f t + h t + bht兲n + 共b f − b兲共h f − 2t兲 h f / 2 + 共b − 2t兲 h / 2 兴 / 关n共2b f t + 2ht + bt兲 + h共b − 2t兲 + 共b f − b兲共b f − 2f兲兴. 2

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Table 4. Evaluation of Load-Carrying Capacity

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Specimens

Vbf 共kN兲

Vbs 共kN兲

DSCW1C — — DSCW2 — — — — DSCW3 共+兲 e — — DSCW3 共⫺兲f DSCW4 543 507 DSCW5 543 507 a V n = M n / l s. b Eq. 共1兲. c doc= ⌺Vnlw / 共M Tn + M Cn兲. d Maximum strength by test. e 共+兲⫽flange wall in compression. f 共⫺兲⫽flange wall in tension.

Vbn 共kN兲 — — — — 507 507

N 共kN兲 0 0 0 0 2,534 2,534

M n共M Tn兲 共kN-m兲 2,715 2,715 3,899 4,415 161 957

M Cn 共kN-m兲 — — — — 663 1,206

sumed that shear yielding developed in all coupling beams, on the basis of the observed failure modes of the wall specimens. As presented in Table 4, the predicted load-carrying capacities correlate well with the test results 共see Vu / Vn in Table 4兲. This result indicates that the assumptions of plastic stress distribution at the cross sections and shear yielding in the coupling beams can be used to predict the load-carrying capacities of the test specimens. Table 4 also shows the degrees of coupling 关=⌺Vnlw / 共M Tn + M Cn兲兴 for two coupled walls. For DSCW4 with two rectangular cross sections and DSCW5 with two T-shaped cross sections, the degrees of coupling were 2.77 and 1.31, respectively, which indicates that 73 and 53% of the flexural moments applied to the specimens were carried by the coupled axial force action of the walls.

ls 共mm兲 3,570 3,570 3,570 3,570 3,720 3,720

4.

lw 共mm兲

Degree of couplingc

— — — — 900 1,122

— — — — 2.77 1.31

Vn 共kN兲 a

760 760a 1,092a 1,237a 835b 1,346b

Vu d 共kN兲

Vu / Vn

869 809 1,291 1,212 822 1,390

1.14 1.06 1.18 0.98 0.98 1.03

prevent the buckling of the steel plates. In the walls with a tie bar spacing-to-steel plate thickness ratio of 30, local buckling of the steel plates was initiated at tensile strains of 0.021– 0.028 mm/mm. The specimens failed at 0.041–0.043 mm/ mm. Despite the local buckling of the steel plates, the plastic stress distribution in the steel and concrete of the cross section could be used to predict the load-carrying capacity of the walls.

Acknowledgments This research was financially supported by POSCO and the writers are grateful for this support.

Summary and Conclusions Five slender double skin composite walls, including isolated walls and coupled walls, were tested to investigate the earthquake resistance of walls subjected to cyclic loading. The test results showed that the double skin composite walls had excellent loadcarrying capacities, but in order to ensure the ductile behavior of the walls, adequate details should be provided to prevent early fracture of the welded connections 共joints兲 at the wall base and coupling beam, and early local buckling of the steel plates. The results of the present study are summarized as follows: 1. Under reversed cyclic loading, double skin composite walls with large depths were susceptible to early fracture of the welded connections at the wall base and coupling beams. This is due to a high concentration of stress at the welded joints, as well as a large plastic strain demand arising from the large depth of the walls 2. In preventing early fracture of the welded connection at the wall base, the cover plate strengthening method, which uniformly increased the steel plate thickness near the connection region, was superior to the rib plate strengthening method. A redundant strengthening scheme, such as the cover plates used in this study, is recommended to make the wall base stronger than the wall. 3. When early tensile fracture in the welded connections was prevented, the walls failed primarily from local buckling of the steel plates, because of the tensile residual strain that developed under cyclic loading. Further, under cyclic loading, the filled concrete did not provide adequate resistance to

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