Composite bridges with prefabricated decks (ELEM)
Research and Innovation
EUR 25897 EN
EUROPEAN COMMISSION Directorate-General for Research and Innovation Directorate G — Industrial Technologies Unit G.5 — Research Fund for Coal and Steel E-mail:
[email protected] [email protected] Contact: RFCS Publications European Commission B-1049 Brussels
European Commission
Research Fund for Coal and Steel Composite bridges with prefabricated decks (ELEM) M. Feldmann, F. Möller, S. Möller
Rheinisch Westfälische Technische Hochschule Aachen (RWTH) Mies-van-der-Rohe-Straße 1, 52074 Aachen, GERMANY
P. Collin, R. Hällmark, O. Kerokoski Ramböll Sverige AB (Ramböll) Box 850, 971 26 Luleå, SWEDEN
M. Kożuch, W. Lorenc, S. Rowiński
Wrocław University of Technology (PWr) WybrzeżeWyspiańskiego 27, 50-370 Wrocław, POLAND
P. Collin, R. Hällmark, M. Nilsson, L. Åström Luleå University of Technology (LTU) Universitetsområdet, Porsön 971 87 Luleå, SWEDEN
B. Norlin
KTH - Royal Institute of Technology (KTH) Brinellvägen 23, 100 44 Stockholm, SWEDEN
G. Seidl, T. Hehne, O. Hoyer, M. Stambuk SSF Ingenieure AG (SSF) Domagkstraße 1a, 80807 München, GERMANY
T. Harju
Rautaruukki Oyj, Ruukki Construction (Ruukki) Harvialantie 420, FI-13300 Hämeenlinna, FINLAND
Grant Agreement RFSR-CT-2008-00039 1 July 2008 to 30 June 2011
Final report Directorate-General for Research and Innovation
2013
EUR 25897 EN
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Table of contents SUMMARY ..............................................................................................................................................................5 SCIENTIFIC AND TECHNICAL DESCRIPTION OF THE RESULTS ........................................................13 OBJECTIVES OF THE PROJECT ...............................................................................................................................13 COMPARISON OF INITIALLY PLANNED ACTIVITIES AND WORK ACCOMPLISHED ....................................................13 1
BRIDGES WITH PREFABRICATED ELEMENTS – STATE OF THE ART .....................................17 1.1 INTRODUCTION ...........................................................................................................................................17 1.2 OVERVIEW ..................................................................................................................................................18 1.3 KNOWLEDGE EXTENSION ACTIVITIES..........................................................................................................20
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CONNECTION SLAB-SLAB......................................................................................................................23 2.1 GENERAL ....................................................................................................................................................23 2.2 LABORATORY TESTS ON POINT TO POINT SHEAR KEYS ................................................................................23 2.2.1 Shear key test ....................................................................................................................................24 2.2.2 Large scale laboratory tests .............................................................................................................31 2.2.3 Conclusions ......................................................................................................................................32 2.3 ALTERNATIVE DESIGN WITH LINEAR SHEAR KEY........................................................................................33
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DURABILITY SLAB-SLAB........................................................................................................................37 3.1 TEST SETUP .................................................................................................................................................37 3.1.1 Test specimens ..................................................................................................................................37 3.1.2 Test rig .............................................................................................................................................39 3.1.3 Measuring equipment .......................................................................................................................40 3.1.4 Climate control .................................................................................................................................41 3.2 TEST PROCEDURE AND RESULTS .................................................................................................................42 3.2.1 Test 1 ................................................................................................................................................42 3.2.2 Test 2 ................................................................................................................................................43 3.2.3 Test 3 ................................................................................................................................................43 3.2.4 Tests 4, 5 and 6 .................................................................................................................................44 3.3 CONCLUSIONS AND DISCUSSION .................................................................................................................44
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SHEAR CONNECTION GIRDER –SLAB ................................................................................................47 4.1 LABORATORY TESTING ON CLOTHOIDAL SHAPED CONCRETE DOWELS ........................................................47 4.1.1 Introduction ......................................................................................................................................47 4.1.2 Steel failure criteria..........................................................................................................................48 4.1.3 Concrete failure criteria ...................................................................................................................49 4.1.4 Tests evaluation and summary .........................................................................................................50 4.2 ADVANCED DESIGN OF SHEAR CONNECTION GIRDER-SLAB .........................................................................52
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EFFECTIVE WIDTH AND STATIC BEHAVIOR ..................................................................................59 5.1 THEORETICAL BACKGROUND ......................................................................................................................59 5.2 LARGE SCALE LABORATORY TESTS .............................................................................................................60 5.2.1 General .............................................................................................................................................60 5.2.2 Test set-ups .......................................................................................................................................61 5.2.3 Results ..............................................................................................................................................63 5.2.4 Analysis ............................................................................................................................................70 5.2.5 Conclusions ......................................................................................................................................77
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FIELD MONITORING................................................................................................................................79 6.1 TEST SET-UP ...............................................................................................................................................79 6.2 RESULTS .....................................................................................................................................................80 6.2.1 Inspection .........................................................................................................................................80 6.2.2 Deflections........................................................................................................................................80
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6.2.3 Steel stresses .....................................................................................................................................81 6.3 ANALYSIS ...................................................................................................................................................82 6.3.1 Deflections........................................................................................................................................83 6.3.2 Stresses .............................................................................................................................................84 6.4 CONCLUSIONS ............................................................................................................................................86 7
DESIGN GUIDANCE AND DESIGN EXAMPLES .................................................................................87
LIST OF TABLES .................................................................................................................................................88 LIST OF FIGURES ...............................................................................................................................................89 REFERENCES .......................................................................................................................................................92 APPENDIX: DESIGN GUIDE DISTRIBUTED IN THE FRAME OF ELEM ................................................95
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SUMMARY The mainn objective of o this projeect is to estaablish a new w type of bridge construcction using full depth prefabricated deck ellements for composite c brridges. Up to o now, comp posite bridge s have reach hed a high level of pprefabricatioon. Neverthelless, typical advantages of prefabrication, like fa fast erection times and high mannufacturing quality, q are partially p com mpensated du uring the on--site assemblling process when the elements are connecteed among themselves wiith an infill concrete. c To shorten the cconstruction time, slab elements with so callled “dry join nts”, i.e. jointts leaving a gap which iss not filled w with concretee between the precaast elements,, are under investigation i n. This type of construction reduces the amount of in situ concrete and hence inncreases the degree of pre refabrication.. Using this technique, t thhe connection n between the slabs and the steeel girder is reealized by a concreting channel c with injection hooles, see Figu ure 1. The channels can be filledd from the to op of the slaab with a fasst hardening infill concreete thus allow wing for a fast consttruction proggress of the superstructur s re.
conccrete chanel
groov ve and tonguee joint
injection / veent holes
Figure 1: Slab elemeents with drry joints To transffer vertical looads betweeen adjacent sslab elementss, groove and tongue joiints are applied which act as shhear keys. Using U these shear keyss, up to 40% % of the wheel w load ccan be transsferred to neighbouuring elemennts. A possible problem tthat arises du uring the ereection is thatt the elements have to be assem mbled with an a initial offfset due to thhe shear key ys before pu ushing the ellements togeether with hydraulicc jacks, see Figure 2. During D this pphase of the erection, the placementt of the stud ds and the transversal reinforcem ment bars in the slab goveern the maxiimum allowaable tolerancees of the pro oduction.
ng process off slab elemeents with sheear keys Figure 2: Assemblin
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There are two possibilities to increase the demanding tolerances: -
the use of a smaller shear key the use of shear connectors which leave bigger distances
Both solutions are under consideration within the project. After the erection of the steel girders, the concrete slabs are assembled and the concrete channels are concreted within a few hours without further need of form- or reinforcement work. Using such type of bridge construction, different questions occur, which are addressed within the project: -
Are there experiences gained in other countries which can be applied for the design of composite bridges with prefabricated deck elements What alternative type of shear connection can be applied between steel girder and slab elements to increase the demanding tolerances How is the bearing behaviour affected by the use of shear keys How can the shear key be designed to decrease the initial offset Is there a composite action over internal support, and How is the sealing to be designed over internal supports, where large gap openings may occur
To answer the first question and to gain knowledge on different construction methods, which are employed in other countries, an international workshop was being held by RAMBÖL in Stockholm (Task 1.1). The main benefits using full depth prefabricated deck elements are a reduced construction time leading to shorter traffic interruptions and a higher quality of the concrete deck among others. These advantages, however, are often compensated when the elements are connected to each other with an infill concrete. To tackle this problem, dry joints were identified as the most promising alternative. The potential problems arising with the dry joint technology were gathered and identified in Task 1.2. A schematic overview over different methods of prefabrication of composite bridges is given in chapter 1 whereas a more detailed overview of the state of art in prefabricated composite bridge design is given in the design guide. To address the second question, the current state of art for the shear transmission between adjacent deck elements was identified and a new solution was elaborated (Task 1.4 and Task 2.2). The depth of the shear key is a crucial parameter in designing a bridge with dry joints, causing a conflict of objectives. On one hand, the shear key should have a sufficient depth in order to transfer loads between adjacent deck elements and thus to reduce the acting bending stress in the longitudinal direction (i.e. the transversal direction of the bridge) of the elements. On the other hand, the size should be as small as possible as the size of a shear key is a limiting factor for tolerances during the assembling process. The size of the tongue directly determines the initial gap between elements and is limited by the distance between shear studs or transversal reinforcement, respectively. To solve these contrasting demands, different test series were performed to describe the failure behaviour of shear keys and to gain knowledge on the main parameters of influence for shear keys. By finite element calculations it could be shown that a maximum of 40% of the traffic load can be transferred by one single shear key. Three test series with a total of 12 tests were performed which focused on the pure shear failure of shear keys in order to proof that shear keys are capable to transfer these loads (Task 3.3). Two series focussed on a different reinforcement lay out, whereas the last series was on unreinforced shear keys. For the reinforced shear keys, two different types of failure could be observed (see Figure 3): -
Cracking of concrete which lead to an activation of the reinforcement (ductile behaviour) Cracking of concrete coverage which lead to a brittle failure of the specimen
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Figure 3: left: ductille failure, right: failure of concrete cover As expeccted, the secoond failure ty ype lead to loower ultimatte loads. Thee tests were eextended by additional large scaale tests withh a smaller sh hear key thaan in the prev vious tests. All A the tests were evaluaated using different calculation models, witth the concllusion that the t shear keeys can be ssafely design ned using formulae according to t Eurocode 2. Furtherm more, it could d be shown that the dessign of the shear s keys proofed a sound desiggn for typicaal loads on coomposite brid dges according to Eurocoode. The assem mbling process is also sttrongly influuenced by thee spacing off shear conneectors. Thus, the shear transmisssion betweenn girder and slab was thhe focus of the t research and new waays of conneecting the girder wiith the slab were w investig gated (Task 11.3 and Task k 2.1). One solution s to inncrease tolerrances and hence to simplify thhe assembling process iss to use con ncrete dowelss. This relattively new method m of connectinng the steel girder g with the t concrete deck elemen nt was under investigatioon in Task 3.1 3 where clothoidaal shaped steeel strips werre tested in ppush out testts in order to o derive desiggn formulaee for ULS. In total 18 POST diviided in 6 testt series were performed. Parameters under u investiigation were -
T The influencee of the web thickness onn the steel ressistance T The influencee of the number of activee dowels on its i resistancee T The influencee of the number of dowell strips and th heir spacing,, respectivelyy
A shape factor AULT of o 0,25 for stteel failure ccould be derived for the ultimate u limiit state. The tests t were also evaluated with respect r to concrete failurre with formulae derived d in the Germ man project P804. P For almost aall test speciimens underr investigati on, concretee failure waas governingg the design n and the formulae lead to a saffe design of the t connectioon. he distance beetween transsversal reinfo orcement is tto use studs in i groups. Another ppossibility too increase th It was iniitially planneed to test thiis alternativee in Task 3.2 2 but as EC 3 already alllows for a deesign with studs in ggroups, anotther critical detail d was exxamined. Ass it is very likely that drry joints inflluence the bearing bbehaviour off the composiite girder, teests were performed to in nvestigate in the effectivee width of compositte girders wiith dry jointss. The tests w were compleemented by non-destructi n ive tests on a segment of a bridge which waas built in laaboratory to substitute a pilot bridge (Task 2.3).. The destrucctive tests focused oon the bearinng behaviour under sagginng moment, see Figure 4. 4
Figure 4: Principal test t set up fo or destructivve tests und der sagging moment m A total oof 4 large sccale tests wiith a length of 9 m and a width of 3.2 m were performed. The tests focused oon: -
T The type of looading (3-po oint and 4-pooint bending)) T The size of thhe gap (≈2 mm m and ≈4 m mm) T The segmentaation comparred to a conttinuous slab
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ments were cast in steeel formworkk what lead to relatively y large initiial gaps bettween the The elem elements. These gapss had to be cllosed during the loading and hence decreased thee initial stiffn ness of the compositte beams. Thhe gap can bee significantlly decreased by using maatch cast elem ments where each new element iis using the previously p casted c elemennt as formwork, meaning g that all eleements must be casted in sequenntial order. This T techniqu ue might leaad to longer concreting times, t but ass the prefabrrication of the slab eelements cann be done ind dependently of the bridg ge constructio on process, tthe time can n easily be compensaated. As a main m outcome of the testss, it can be concluded c th hat two diffeerent states have h to be distinguisshed, one whhere the gap ps are open aand the otherr where the gaps are alre ready closed.. The first state migght develop leess distinct for f small gapps, where the gap closes very v fast probbably even under u selfload of thhe bridge strructure. Anyh how, the stifffness of the composite beam b decreasses and lies below the stiffness of a girder with w full com mposite actioon. A recom mmendation for f the calcuulation of thee effective width forr displacement calculations is provideed as an alterrnative for fiinite elementt calculations. It could also be shown withhin the tests, that the uultimate bearring capacity is affecteed only sligh htly by a segmentaation of the slab s and that the test speccimens could d reach the ultimate u bearring capacity y using the effective with accordding to EC 4. Special atteention has to be paid to th he size of ini nitial gap: alth hough the test speciimens reachhed a sufficieent bearing capacity, thee specimens failed by a crushing off concrete coverage due to highh stress concentrations inn the contactt area betweeen adjacent eelements. Th he risk for this failurre can be deecreased by reducing r the gap size, ho owever, if it is likely that at the joint gaap will be big > 0,55 mm, it is strongly s reco ommended too perform a non-linear FE-analysis F ssimulating th he gaps in the jointss that are clossing under an n increasing load. In the non-destructivee tests, two different d test set ups weree used to testt the test speecimen in hogging and sagging m moment, see Figure 5.
Figure 5: Test set up p non-destru uctive tests, left: hoggin ng, right: sag gging momeent w loaded only o below the failure load to test the differennt test set ups. u After The test specimen was oaded until failure f to testt the ultimatte bearing performinng the non-ddestructive teests, the lastt test was lo load of thhe shear key.. The test speecimen had a length of 7..4 m and a width w of 3.5 m m, see Figuree 6.
ns of the testt specimen Figure 6: Dimension
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The tests focused on: -
P Possible com mposite actio on under hoggging momeent regarding g deflectionns, joint open nings and sstresses A Amount of foorce which iss introduced in the reinfo orcement und der hogging m moment C Composite acction in areass of sagging moment C Change in beehaviour due to load cyclees E Effective widdth in areas of o sagging m moment
Within thhe non-destruuctive tests, the t prefabriccated elemen nts were matcch cast with ggaps of approximately 0.5 mm. Besides num merous meaasurements oof strains an nd deflection ns, the testss were simu ulated and comparedd with finite element calcculations, seee Figure 7.
ment model of the non-d destructive tests t Figure 7: Finite elem ment, the testts could con nfirm the resu ults of the ddestructive teests in the For the ttests under sagging mom followingg points: -
U Under sagginng moment, the t overall sttiffness decreeases A After closingg of the gap ps under higgher load, th he girder accts as a com mposite bridg ge with a ccontinuous sllab If the gaps arre too big, a finite elemeent calculatio on should bee performed tto study the influence oof the gaps onn the stressess and deflecttions
The testss under hoggging bending g moment shhowed that th he influence of the concrrete is very small and for the deesign in the ULS U only thee steel cross section shou uld be taken into i account.. To providde alternativves for the afforementioneed problems, occurring during the asssembling pro ocess, two innovativve bridge dessigns were ellaborated by SSF Ingenieeure in Task k 2.2, which ccould both be b realized during thhe runtime off the project: -
A bridge usinng prefabricaated elementss as a compo osite grillage (called VTR R Technology y) A bridge witthout compo osite action bbetween steeel girder and d prefabricatted concrete elements uusing a linearr shear key
The VTR R bridge, see the left picture of Figure 8, takes t advanttage of the prefabricatiion using prefabricated concrette modules but b allows foor higher toleerances due to t the open jjoints which h are filled with concrete. This measure m lead ds to a goodd compromise between a short constrruction timee and high possible tolerances. The second bridge typee being desiigned within n the projectt is a bridgee with no compositte action andd dry joints between b preccast elements. In this bridge type, noo concrete works w have to be carrried out onn the constru uction site. The linear shear key, however, h neecessitated very v small tolerancees to fit oveer the wholle bridge wiidth. Thereffore, the sheear key hass to be grin nded after concretinng.
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Figure 8: Left: VTR R bridge; rig ght: prefabriicated deck element witth linear sheear key Whilst thhe latter menttioned detaills affect the bbridge in its bearing b behaaviour, the laast critical deetail under investigaation is of coonstructional nature. By cconnecting th he deck elem ments only byy means of dry d joints, high straiins can occuur in the insu ulating layer in the area of internal support wherre the gaps open. o This detail waas under inveestigation in n Task 3.4 w where fatiguee tests on asphalt and innsulation layeers of dry joints weere performedd in laboratory, see Figurre 9.
Figure 9: Fatigue tessts on aspha alt layer mperature of -20°C. In aaddition to th he load in The testss were perfoormed under wet conditioons at a tem longitudiinal directionn of the speccimen, whichh opened an nd closed thee gap, a verttical load waas applied that simuulated an axxle load of a truck. In ttotal 6 cycliic tests weree performed on 4 differrent water proofingss. The main outcome of these t tests iss that it is nott possible to allow for anny substantiaal deck-todeck opeening if the water proofing membrranes are jusst welded to o the deck, and that an n artificial debondinng zone by means m of e.g.. PTFE strip s has to be applied a in orrder to distribbute the elon ngation of the membbranes over a greater len ngth. A princcipally good fatigue desig gn that withsstood over 4.5 million cycles is depicted in Figure F 10.
Figure 10: Water prroofing with h debonding zone
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a the dry joint tecchnology forr a bridge witith an internaal support, As no new bridge couuld be built applying an existinng one spann bridge wass monitored, which was built ten yeears ago usinng prefabricated deck elements and dry jooints (Task 4.1). The bbridge was loaded by a truck sim milar to a monitoring m programm me which waas carried ou ut 10 years beefore, see Fig gure 11.
Figure 11: Monitoriing of a 10 year old brid dge with preefabricated decks d and ddry joints with the resu ult that no In additioon the bridgge was inspected by the Swedish roaad administraation STA w damages have occurrred which are a related too this speciaal bridge typ pe. The meaasurement prrogramme compriseed strain meeasurements on the steeel girder in three differeent sections and three deflection measurem ments on eaach steel girrder. The m measurementss were comp pared with the results from two different calculation methods: m -
U Using a simpplified beam model m whichh was initiallly applied forr the design oof the bridgee U Using finite element e metthods, three ddifferent mo odels were seet up with diifferent apprroaches to m model the back walls and d the soil-to-bback-wall intteraction
The majoor outcome of o the monittoring is thatt the measurred results off the two tesst campaignss show no significannt differencees and that th he measuredd results corrresponded veery well with th the resultss from the finite elem ment calculaations. Furthermore, the rresults were compared with w the new gained know wledge by laboratorry testing andd complemen nt the results of Tasks 3.3 3 and 3.2. The outccome of the project is a design guidde with desiign exampless (Task 5.1 and 5.2). The design guide coontains desiggn rules derrived from tthe experim mental campaaigns, design gn and manu ufacturing recommeendations gaiined by the experience e oof RAMBÖL LL and work ked exampless for critical details of compositte bridges wiith prefabricaated deck eleements and dry d joints.
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SCIENTIFIC AND TECHNICAL DESCRIPTION OF THE RESULTS Objectives of the project The overall objective is to improve competitiveness of composite bridges by an elaboration of a new cost effective, time effective and sustainable bridge structures. To allow for a prefabrication of the concrete deck, a new concept with dry joints was investigated and developed.
Comparison of initially planned activities and work accomplished To reach the objectives stated above, the state of the art in prefabricated bridge design was identified and typical solutions for different construction details were reviewed. By a catalogue of critical detail for the bridge type under investigation new solutions were elaborated and tested. These solutions cover the load transfer between adjacent slabs, the load transfer between girder and slab, the overall bearing behaviour and the fatigue resistance of the insulation layer over internal support. A monitoring of an existing bridge round out the performed work. WP 1 “Economic Analysis and design evaluation” was completed and a Workshop was being held in Stockholm. The proceedings of this workshop can be downloaded at: http://pure.ltu.se/ws/fbspretrieve/3112371. Different designs were developed in WP 2 “Advanced design of match-cast-prefabricated slab element bridges, dry joints” for the critical details identified in WP 1. A drawback, however, can be seen in the fact that it was not possible to find any contractor for a multi-span pilot bridge. Therefore, it was decided to substitute the pilot bridge with expanded laboratory testing, which is described in Task 3.3. WP 3 “Tests on match-cast-prefabricated slab element bridges” comprises 4 different Task which will be described separately. Task 3.1 “Durability of girder – slab connection: concrete dowel” was mainly performed at Wrozław University. Instead of a combined connection comprising concrete dowels and conventional studs, a clothoidal shaped concrete dowel was under investigation, for which no design formulae were derived so far. In sum, 18 static push out tests according to EC 4 were performed as promised. Task 3.2 “Capacity of girder – slab connection”: As it turned out that studs in groups have been studies in detail in the past and that EC 4 allows for a design of studs in groups, it was decided to study the bearing behaviour of composite beams with dry joints. Of special interest was the question, if the effective width can be calculated according to Eurocode for such structures and if the ultimate load is reached. To answer these questions large-scale 3 and 4 point bending tests were performed. The tests were performed at RWTH Aachen University until ultimate failure so that knowledge could be gained on the effective width and the ultimate bearing capacity. Furthermore, two tests under negative bending were performed. These tests should answer the question if the reinforcement of the slab element could be taken into account over internal support. The result of these tests was that the load cannot be transferred into the concrete within one slab element and that the reinforcement should not be taken into account over internal support. Task 3.3 “Capacity of slab – slab connection” was executed by Lulea University and Ramböll. Overall 12 tests on shear keys were promised within the proposal. Two tests series were performed. The first tests series focused on a pure shear failure in the concrete shear key. Twelve static tests with three different layouts of the shear keys were tested. The second test series was a large scale laboratory tests as a substitution for the pilot bridge. In this series, the main purpose was to study the large scale behaviour of a prefabricated bridge deck with dry joints. The bridge segment was tested under different
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loading pattern and investigated the bearing behaviour such as the effective width. But some of the tests were focused on the capacity of the shear keys and the distribution of forces between the different rebars inside the shear key. In Task 3.4 “Durability of water proofing of open joints” 1 cyclic test should be performed initially under wet conditions. As bitumen is very temperature dependent, it was decided to perform the tests at a temperature of -20°C in a climate chamber. Overall, six tests were performed with 4 different set ups of the insulation layer. WP 4 “Monitoring” was performed on an 11 year old single span bridge, which was built using the dry joint technology. During the monitoring deflections and strains were measured. Parallel to the monitoring, the bridge was inspected by the Swedish road authority in order to assess the condition of the bridge after 11 years of usage. The results were evaluated and compared to finite element and analytical calculations as promised. Table 1 gives an overview of the work performed within the project.
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Table 1 Programme bar chart ELEM Time Schedule 2008 finished WP 1.1
I
II
III
2009 IV
I
II
III
2010 IV
I
II
III
2011 IV
I
II
III
IV
EXTENSION OF KNOWLEDGE (WP Leader: Ramböll) 100 %
WP 1.2
CRITICAL DETAILS (WP Leader: KTH) 100 %
WP 1.3
SHEAR TRANSMISSION GIRDER-SLAB (WP Leader: SSF) 100 %
WP 1.4
SHEAR TRANSMISSION SLAB-SLAB (WP Leader: Ramböll) 100 %
WP 2.1
IMPROVED GIRDER-SLAB (WP Leader: SSF) 100 %
WP 2.2
IMPROVED SLAB-SLAB (WP Leader: Ramböll) 100 %
WP 2.3
PILOT BRIDGE (WP Leader: Ramböll) 100 %
WP 3.1
DURABILITY GIRDER-SLAB, CONCRETE DOWEL (WP Leader: SSF) 100 %
WP 3.2
CAPACITY GIRDER-SLAB, SHEAR CONNECTORS IN GROUPS (WP Leader: RWTH) 100 %
WP 3.3
CAPACITY SLAB-SLAB (WP Leader: LTU) 100 %
WP 3.4
DURABILITY OF WATER-PROOFING OF OPEN JOINT (WP Leader: KTH) 100 %
WP 4.1
MONITORING (WP Leader: LTU) 100 %
WP 5.1
DESIGN GUIDANCE (WP Leader: RWTH) 100 %
WP 5.1
DESIGN EXAMPLES (WP Leader: RWTH) 100 %
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1 BRIDGES WITH PREFABRICATED ELEMENTS – STATE OF THE ART 1.1 Introduction Determining the most efficient and economical way to build a new or replacement bridge is not an as straightforward process as it once was. The total cost of a bridge project is not limited to the amount spent on concrete, steel and labour. Construction activities disrupt the typical flow of traffic around the project and results in additional costs to the public by means of longer waiting times, additional mileage travelled to bypass the construction site, or business lost due to customers avoiding the shops affected by the construction site. The risk of injury to workers due to traffic interactions or construction activities increase with each hour spent at the construction site. Finding a way to shorten the time spent on the job site is beneficial to the Contractor, the Owner, and the travelling public. In recent years, transportation agencies around the globe have begun to use Accelerated Construction Techniques (ACT) which incorporate Prefabricated Bridge Elements (PBE). Prefabricating certain bridge elements reduces the time spent at the construction site and reduces the impacts on the road users and the surrounding community. As an example, steel beams with composite concrete decks reduce the construction time over cast-in-place concrete structures. Additional time savings can be achieved by prefabricating the riding deck and the substructure. In some instances, entire structures have been fabricated off-site under strict environmental and quality controls and then shipped to the site and erected in a few days instead of months. Unfortunately, ‘road user costs’ are often neglected when design alternatives are compared. The total expense of using prefabricated bridge elements is usually comparable to traditional construction techniques. However, the overall cost may be significantly less than traditional construction when durability and ‘user costs’ are taken into account (Culmo, 2009). The total cost of using prefabricated bridge elements depends greatly on the scale of the prefabrication. The more times that prefabrication is repeated the lower the costs. A U.S. study (Ralls 2008) states that, for agencies that use prefabrication infrequently, the initial cost of a structure built using PBE is slightly higher than for traditional piecemeal construction procedures. However, it is the overall, or ‘life-cycle’ costs of PBE structures that must be compared to traditional construction to determine which provides the best value. Transport agencies in USA that use prefabrication more consistently find that project bid prices are in line with, or sometimes lower than, traditional construction as Contractors become more familiar with the methods. To capitalize and build upon this experience, the Massachusetts Department of Transportation (MassDOT) has begun to ‘bundle’ multiple Accelerated Bridge Construction projects along a travel corridor into a single contract (MASSDOT Website, 2011). This way, the designs on similar projects can be produced more efficiently and the construction becomes repetitive and more familiar. ‘Bundled’ design projects, as opposed to having separate contracts for each project, save procurement time and leverage the Contractor’s expertise and momentum. MassDOT has already solicited bids on thirteen bundling contracts that will address repairs to several hundred substructures and bridge decks, primarily located on interstate highways and major arterials across the Commonwealth. A cost-benefit analysis (Degerman, 2002) of a single span railway crossing bridge in Norrfors, Sweden that used PBE showed a significant reduction of costs and construction time. The cost to prefabricate the bridge was presumed by the designers to be higher than the cost to cast the concrete deck on site. However, a bidding Contractor indicated that the prefabricated deck was actually less expensive, as the additional costs due to prefabrication is overcome by the savings achieved by eliminating formwork and other activities required when building a bridge over an operating railway. A study of a Swedish road bridge (Nilsson 2001), called the Rokan Bridge, compared the cost of constructing a prefabricated composite bridge, a conventional composite bridge with a temporary detour bridge, and a conventional composite bridge with the traffic directed to the nearest bypass roads. The
17
conclusion from the study was that the most economical alternative, from a societal perspective, was to construct a prefabricated bridge and eliminate the need for the long-term detours. Even where there are advantages to using prefabrication, typical construction techniques are still often used out of familiarity and habit. Going forward, prefabricated bridges should consistently result in lower initial construction costs and produce final products of higher quality once the prefabricated elements, connection details, construction procedures, and other details are standardized and become more familiar.
1.2 Overview The composite construction has gained a high level of prefabrication in order to reduce construction time and hence increase competitiveness. Recent developments lead to a prefabrication of the concrete slab where different construction methods can be distinguished:
Level of prefabrication
on site concrete
half depth prefabricated concrete elements
full depth prefabricated concrete elements
Figure 12: Level of prefabrication The main field of application for half depth prefabricated concrete elements are bridges with small spans where the use formwork construction is expensive due to the small span (Geißler 2009). The partly prefabricated elements are assembled on the steel girder and are used as an integrated formwork. They contain starter bars for further concreting. Full depth prefabricated concrete elements differ from the latter technique as no additional in situ concrete is being used. This technique is very robust and the resulting concrete slab is of very high quality. However, the execution of the joints between concrete elements and between concrete elements and steel girder is a critical detail. Table 2 and Table 3 give an overview of such joints.
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Table 2: Possible typ pes of longittudinal jointts open joinnts
The elem ments are laid on the steeel structure aand connecteed by reinfoorcement barrs. + easy too assemble - possibiility of crack ks between in n situ concreete and precaast elementss - slow w work progreess as in situ u concrete hhas to hardeen before thhe work on th he pavement can start
pockets
The studds can be weelded on the steel girder in the shop or on the coonstruction site. The conccrete deck haas pockets.
Ausspaarungen
+ less cooncrete necesssary + the preecast elementts can be big gger - studs caan interact
niches
The slabb has holes with a diam meter of 80 mm in whicch studs aree welded on site s after positioning of thhe slab.
Nischen Ø 80
+ the stuuds can be eq qually distribu uted in the sllab Fuugenband
p + no probblems with positioning - studs arre welded in the holes on n site
channels Querbeweh hrung
The slabb has a chann nel with tran nsverse reinfforcement an nd injectionn holes.
Einfülllocch
+ fast haardening mo ortar can be used which leads to veery short connstruction tim mes Kopfbollzendübel
- high deemands on to olerances encased ssteel plate
A steel pplate is encassed in the con ncrete slab. T The steel plaate is weldedd on the steel girder.
Zementinjektion
+ the c omposite acction betweeen steel and nd concrete is establishhed in the sho op Baustellennah ht
- high am mount of wellds on site - the wellds cannot beear transversaal bending m moments
no compoosite action
The slab is laid upon n the steel girder. + easy too install - no com mposite action n
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Table 3: Different types of transversal joints Wet joints using concrete
Wet joints using cement mortar
injection joints using epoxy mortar
Dry joints
Thickness
≥ 100 mm
≤ 30 mm
< 3 mm
≈ 0 mm
tensile coupling between elements with reinforcement
possible
not possible
not possible
not possible
hardening time
long
long
fast
none
compensation of unevenness
possible
possible
partly possible
not possible
temporary prestressing necessary
remarks
1.3 Knowledge extension activities An international workshop was organised by Ramböll providing a forum for experts for the presentation and discussion of their experiences within the field of element bridges. The workshop took place in Stockholm, Sweden, on 4th March 2009. The workshop was free to attend, the number of attendants was however restricted to 50 persons in order to get good conditions for discussions. All of the partners in the RFCS-project were invited together with interested persons in their contact network. A public invitation was published on two separated homepages, the homepages of Ramböll and the Swedish Steel Institute. The later also distributed the invitation by e-mail to their contact list, including people from different part of the construction business. An advertisement was also published in a Swedish construction journal. Beside the public invitations, personal invitations were also sent to key persons. A couple of international experts were also invited as key speakers. The speakers were asked to present experiences from prefabricated bridges in their home countries. The agenda of the workshop, including speakers and topics, is given in the table below. Six different countries were represented among the speakers. Table 4 Programme of the ELEM-workshop The need for cost- and time effective Ingemar Skogö, Swedish National infrastructures Road Administration Composite element bridges in Sweden Peter Collin, Ramböll/LTU Prefabricated composite bridges in the US Mike Culmo, USA total bridge prefabrication and installation French experiences from element bridges Jacques Berthellemy, SETRA Composite Element bridges in the UK Stuart Gordon, UK Composite Element bridges in Germany Günter Seidl, SSF Ingenieure Timo Tirkonen, Finnish Road Composite Elment bridges in Finland Administration RFCS project ELEM Daniel Pak, RWTH Aachen
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Participants 55 persons attended the workshop. There were participants from 9 different countries, Sweden, Finland, Denmark, Germany, France, USA, UK, Poland and Estonia. The companies, universities and authorities represented are shown below. Bridge design companies:
Ramböll, WSP, ELU, Grontmij, URS Corpo, Tyrens, SSF-partners, CME Associates.
Universities
Luleå University of Technology, RWTH Aachen, KTH Stockholm, Wroclaw University of Technology. Watts University Edingburgh.
Road Authorities
Swedish National Road Administraion, Finnish Road Administration, SL-Stockholm. SETRA France.
Manufacturers
Ruukki, Strängbetong, Outokumpo Stainless.
Contractors
Skanska, PEAB, Prefabsystem, NCC.
Outcome A technical report has been published, summarizing the presentations held at the workshop: Collin, P.,Hällmark, R., Nilsson, M: “International Workshop on Prefabricated Composite Bridges”, Luleå University of Technology, Lulea, 2009 (ISBN: 978-97-7439)
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2 CO ONNECTIO ON SLAB-SLAB 2.1 G General The studyy of the slabb-slab connecction has beeen focused on o prefabricaated deck syystems with dry d joints. To transfer both latteral and veertical forcess through th he transverse joints, and nd to preven nt vertical displacem ments betweeen the deck elements e at tthe joints, ov verlapping co oncrete keyss are used. These T keys are eitherr designed ass a series of overlapping male-femalee connection ns (point to ppoint shear keys) k or as a linear sshear key aloong the comp plete width oof the joints. In order to avoid miss m match in the dry joint, the elemeents can be match-cast m orr grinded afteer concreting g. Hence, tw wo different types of sheaar keys weree investigated d in within th he research pproject: -
P Point to pointt shear keys at Lulea Uniiversity
-
L Linear shear keys at SSF Ingenieure
The abiliity of the poiint to point keys k to carryy shear forcees has been investigated, i , keeping in mind that the elemeents are partss of a compo osite structurre, i.e. they cannot c separaate since theyy are kept to ogether by the steel girders. A design mod del of dry cooncrete jointts, modelled d as compresssive struts, has been supportedd by extensivve testing as well as theorretical studiees of the failu ure criterion of such join nts. If possiblle, it would be preferable to use sheaar keys with h smaller dep pth. Howeverr, the shear keys k must be able too transfer thee forces giveen in the desiign codes (E EN 1991-2). By B using a ssimplified FE E-model it can be shhown that a maximum m off about 40% of the trafficc load acting g on a single element is trransferred through oone of the jooints, see Fig gure 13, assuuming a gird der spacing of o 5,0 m andd an element length of 1,8 m. T The rest of thhe load is transferred dirrectly to thee steel girderrs, or througgh the dry jo oint at the opposite side of the element. e Theerefore, the sshear keys must m be able to t resist a loaad that is at least 40% of the traaffic loads givven in the co odes. The loaad distributio on between th he shear keyys is dependent of steel girders sppacing, num mber of shearr key, the leength of the element, vaarying thicknness of deck k slab etc. Thereforee, it is necesssary to alway ys check thiss value when n the structure is changedd.
Figure 13: Example from FE-m model for loaad transfer through t concrete tonguees (40%).
2.2 Laboratorry tests on o point tto point shear s key ys The first tests series in the laboraatory have bbeen focused d on a pure shear s failure in the concrete shear key. Tweelve static tests with threee different llayouts of the shear keyss have been ttested. The test t set-up and the sppecimens aree briefly desccribed in thee following seections.
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I this case, the t main purrpose was to study the The second test series was a largee scale laborratory tests. In B some of tthe tests werre focused large scalle behaviourr of a prefabrricated bridgge deck with dry joints. But on the caapacity of thhe shear keyss and the disstribution of forces between the diffeerent rebars inside the shear keyy.
2.2.1 S Shear key y test Test set-u up The testss were focussed on pure shear capaccity of the concrete keyss. This meanns that no positive p or negative effects weree simulated, such as presstressing from m the steel girders, g or anny misfit between the elements. A schematiic and simpliified sketch oof the test sett up is shown n in Figure 114.
up Figure 14: Test set-u Test speccimens The geneeral geometryy of the test specimens s w were 1,8 x 1,3 3 m, with a concrete c sheaar key depth of 60 mm and a lenngth of 540 mm, m see Figu ure 15. The sspecimens were cast in a concrete witth a cube tesst capacity that shouuld equals the t strength h class C30//37. For eaach specimen n, six concr crete cube teests were performeed, 3 compresssive and 3 tensile. t
Figure 15: General geometry g off test specim mens.
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The firstt specimens were reinfo orced with exactly thhe same am mount of rein nforcement as used in deck ellements in previously p constructted single sppan bridges. In these specimenns the shear keys k were th he same in both endds. The secoond type of specimens had redduced shearr key reinfforcement, comparedd to the firstt specimens, in one of the shearr keys. The second sheaar key (in specimenn type 2) waas completelly without reinforcement. Withh this desig gn, 4 test results iss gain for eaach type of shear key. Figure 166 shows the reinforcemen r nt drawing of the seccond type off specimens. Figure 16: Reinforcem ment drawingg, specimen 2. Results Two diffferent kinds of o failures were w observedd when the reinforced r sh hear keys weere tested. Fiirstly, five of eight sspecimens faailed by crack ks that activaated the reinfforcement, giving g a ductiile behaviour – failure type 1. The shear keyys remained as a one piece,, but with som me concrete crushing in the lower paarts. Three specimenn failed by cracks c that were w developped outside the reinforccement, resuulting in a faailure that separatedd the shear key k from thee rest of the specimen – failure typee 2. This typpe of failuree occurred under low wer loads thaan the previously describeed failure. d (4 tests) Shear keyy type 1 – Ø112 reinforced Two of ffour shear keeys of type 1 resulted in failure type 1. The load d-time curvess from thesee two tests are shownn in Figure 17 1 with solid d lines, togethher with som me photos of the t failed she hear keys, Fig gure 18.
V [kN]
600 500
100 03 15-1 1
400
100 03 16-2 2
300 200 100 0 0
200
400
600 0 Time [s]
80 00
1000
Figure 17: Load-tim me curve for shear key tyype 1, failurre type 1 (solid) and 2 (ddashed).
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Figure 18: Shear key type 1 with a failure activating the reinforcement The last two specimens failed by cracks that developed outside the reinforcement, resulting in a failure that separated the shear key from the rest of the specimen. The load-time curves from these two tests are shown with dashed lines in Figure 10, together with some photos of the failed shear keys, see Figure 19.
Figure 19: Shear key type 1 after failure in the concrete covering layer. Shear key type 2 – Ø8 reinforced (4 tests) The load-time curves from the tests are shown below together with some photos of the failed shear keys, see Figure 20. In one aspect, the results from these tests reminds of the tests of shear key type 1 – Ø12 reinforced, since three shear keys remains rather unaffected after cracking, and one shear key failed outside the reinforcement bars. 500 100407
450
100318
400
100408 100412
350
V [kN]
300 250 200 150 100 50 0 0
200
400
600
800
1000
Time [s]
Figure 20: Load-time curve for shear keys of type 2. Shear key type 3 – unreinforced (4 tests) The load-time curves from the tests are shown below in Figure 21.
26
1200
160 100407-oarm
140
100318-oarm 100408-oarm
120
100412-oarm
V [kN]
100 80 60 40 20 0 0
50
100
150
200
250
300
350
400
450
500
Time [s]
Figure 21: Load-time curve for shear keys of type 3. For each specimen, six cubes were cast out of the same concrete mix. Three of the cubes were used to determine the compressive strength and the other three were used to determine the tensile strength. The test cubes had the dimensions of 150×150×150 mm. The mean values for each specimen are presented in Table 5 below. Table 5: Concrete parameters Cast date
Test date
Age
Pc
fc
Pct
fct
[days]
[kg/m³]
[kN]
[MPa]
[kN]
[MPa]
2010-03-15
2010-06-16
93
2334
1045
46.1
118
2.6
2010-03-16
2010-06-11
87
2345
1132
49.7
123
2.8
2010-03-18
2010-06-02
76
2372
1082
47.6
103
2.3
2010-04-07
2010-06-08
62
2330
967
42.6
94
2.1
2010-04-08
2010-06-11
64
2358
1009
44.5
115
2.6
2010-04-12
2010-05-31
49
2371
970
42.9
99
2.3
Analysis The strength of the shear keys has been estimated by using four different design models: 1. classic beam analysis – shear key type 3 with failure type 2 2. Eurocode 2 – shear key type 1 and 2 with failure type 2 (EN 1992-1-1) 3. Eurocode 2 – shear key 1 and 2 with failure type 1 (EN 1992-1-1) 4. a force equilibrium model The material parameters from the concrete cube tests are used to calculate the shear capacity for each shear key. Results from the design models are presented in Table 6. 1. Shear resistance for shear key type 3 with failure type 2 The shear resistance for failure type 2 in shear key type 3 without reinforcement can be estimated as, according to classic beam analysis and assuming the shear strength is half the tensile strength (Betonghandboken 1990), 2 1 VRd ,c bw hf shear bw hf ct 3 3
( 1)
where bw = 540 mm; the smallest width of the cross-section within the effective height
27
h fct
= 165 mm; height of shear key = is the tensile strength of the concrete, see Table 1
2. Shear resistance for shear key type 1 and 2 with failure type 2, The shear resistance for failure type 2 can also be estimated by using formulas from (BBK04, 2004),
Vc bw df v
( 2)
f v 0.3 (1 50 ) f ct
( 3)
As 0 /(bw d ) 0,02
( 4)
where ξ d bw fv fct As0
= 1.4 for d ≤ 0.2 m = 140 mm; effective height = 540 mm; the smallest width of the cross-section within the effective height shear strength of the concrete tensile strength of the concrete the smallest amount of bending reinforcement in the tensile part of the studied cross-section. This is set to 0, since there is no bending reinforcement in the shear key, only shear reinforcement.
3. Shear resistance for shear key 1 and 2 with failure type 1, This approach has been used on at least two bridges in Sweden, a bridge over Rokån and a bridge in Norrfors. According to EN 1992-1-1, shear resistance for a section with inclined shear reinforcement can be modelled as A VRd ,s sw zf ywd cot cot sin ( 5) s When shear reinforcement are used locally, with inclined rebars in one line (the –SX rebars), then the equation above can be simplified to
VRd ,s Asw f ywd sin
( 6)
where Asw fyw α
area of the shear reinforcement = 500 MPa; yield strength of the shear reinforcement angle of the shear crack (45° observed in the test) = 60°; inclination of the shear reinforcement
4. Force equilibrium model. This model has been suggested, by Dr. Bo Westerberg (KTH, Stockholm), in order to describe the load carrying capacity in more detail. It is a force equilibrium model that involves both the reinforcement and compressive struts in the concrete, see Figure 22.
28
uilibrium model m and nootations Figure 22: Force equ ved to have great influen nce on the load l carryingg capacity. Without W a The horizzontal force H is believ compresssive horizonttal force, theere is a risk for shear faailure of the concrete covver at the ed dge of the shear keyy. The load carrying c cap pacity is hardd to predict in i such a sceenario, but itt cannot be more m than the shearr strength of the t concrete.. nt on the shaape of the suupports, the rigidity r of In the labboratory tests, the size off the force H is dependen the test-riig etc. On a real r bridge, this t force wi ll vary along g the bridge and a will depeend on the global load situation as well as thhe local load situation. Equilibriuum equationns:
F : V F2 cos F3 0 F : H Fc F2 sin F1 0 M : Vc F1b H b a Fc b z 0
( 7) ( 8) ( 9)
The loadd carrying caapacity of thee shear key ccan be estim mated by usin ng the maxim mum capacitty of each rebar.
F1 f y As1 cos
( 10)
F2 f y As 2
( 11)
F3 f y As3
( 12)
giving mm Ø2 = 12 m F1, max 5500 2 F2, max 5500 8 F3, max 5500 3
Ø2 = 8 mm
16 2 4
12 2 4
12 2 4
12 2
cos18.5 191 kN
F1, max 500 2
452 kN
F2, max 500 5 8
170 kN
F3, max 500 3
4
82 4
12 2 4
cos 18.5 107 kN
201 kN 170 kN
As a firstt assumptionn the horizonttal force, H, is set equal to zero. Then n we assumee that we aree using the shear reinnforcement up u to 100%. This gives thhe following result by equ uation ( 7).
Vmax 4552 cos 30 170 561 kN N
Vmax 20 01 cos 30 170 1 344 kN N
29
The moment equilibrium equation ( 9) gives: Fc
561 66 191 65 582 kN 65 20
344 66 107 65 349 kN 65 20
Fc
Assuming that the compressive strut in the concrete is developed over a height of 30 mm and the width of 540 mm, the compressive stress in the concrete can be calculated as: Fc = Fc/(wh)
Fc = Fc/(wh)
Fc = 582/(54030) = 35.9 MPa
Fc = 353/(54030) = 21.8 MPa
These compressive stresses are below the compressive strengths that have been measured, and failures caused by concrete crushing could not be observed in the tests. The result above would be a possible solution according to this load model, resulting in yielding in the shear reinforcement. Anyhow, this is only one possible solution for this model, based on theoretical positions of the rebars. This load model must be calibrated to the test results, and the influence of the horizontal force H must be investigated. For example, frictional forces between the concrete surfaces will influence the result. Test results vs. calculation models Table 6: Test results compared to results from calculation models.
Cast date
Test results
Model 1
Model 2
Model 3
Model 4
Vmax [kN]
Vmax [kN]
Vmax [kN]
Vmax [kN]
Vmax [kN]
2010-
Ø12
Ø8
-
Ø12
Ø8
-
*
Ø12
Ø8
-
*
Ø12
Ø8
-
*
Ø12
Ø8
-
*
03-15
449
-
-
80
-
-
5.61
84
-
-
5.36
392
-
-
1.15
561
-
-
0.80
03-15
337
-
-
80
-
-
4.21
84
-
-
4.03
392
-
-
0.86
561
-
-
0.60
03-16
532
-
-
83
-
-
6.41
88
-
-
6.07
392
-
-
1.36
561
-
-
0.95
03-16
370
-
-
86
-
-
4.30
88
-
-
4.22
392
-
-
0.94
561
-
-
0.66
03-18
-
285
-
-
68
-
4.19
-
73
-
3.88
-
174
-
1.64
-
344
-
0.83
03-18
-
-
104
-
-
68
1.53
-
-
73
1.42
-
-
0
-
-
-
0
-
04-07
-
222
-
-
63
-
3.52
-
67
-
3.30
-
174
-
1.28
-
344
-
0.65
04-07
-
-
114
-
-
63
1.81
-
-
67
1.70
-
-
0
-
-
-
0
-
04-08
-
363
-
-
77
-
4.71
-
82
-
4.42
-
174
-
2.09
-
344
-
1.06
04-08
-
-
123
-
-
77
1.60
-
-
82
1.50
-
-
0
-
-
-
0
-
04-12
-
376
-
-
68
-
5.53
-
71
-
5.30
-
174
-
2.16
-
344
-
1.09
04-12
-
-
82
-
-
68
1.21
-
-
71
1.16
-
-
0
-
-
-
0
-
* = test result divided by the predicted value for the given calculation model.
According to the result presented in Table 6, calculation model 1 and 2 can be useful to estimate the strength for a shear key without reinforcement. The design values are on the safe side with a safety factor from 1.16 – 1.81. Model 3 gives results that are on safe side except for the failures in the concrete covering for test specimen type 1. With the assumptions made, design model 4 is the same as model 3, except the fact that model 4 makes the vertical reinforcement bars in the slab active. The result is often on the unsafe side, which could indicate that the vertical rebars does not influence the load carrying capacity as much as assumed in the calculations. This model needs to be studied more detailed, calibrated to the test results and maybe modified. One thing that can be noted is that the shear keys that fail in the concrete covering layer still transfer forces that are far higher than the capacity of the concrete itself. Therefore, the reinforcement must have been activated, and should be included in the design formula in one way or another.
30
Non-lineear FE-Analy lysis A detaileed non-linearr FE-analysiss have been ddone for the unreinforced d shear keyss (by Mikael Hallgren, KTH), inn order to coompare the ultimate faillure load fro om FE-analy ysis and the test results.. The FEanalysis’ have been done d in the FE E-software A Atena 3D, an nd with real material m paraameters that have h been tested in the laboratoory. Figure 23 2 below illuustrates the FE-model. F Since S the testted element is doubly symmetriic, only a quuarter of thee element is modelled, with w boundarry condition equalling a complete element.
Figure 23: FE-modeel of shear key tests. Shear keyy type 3 – noo reinforcemeent The FE-aanalysis givees an ultimaate failure looad of P = 124 kN, and a post failuure capacity of 99 kN. Figure 244 below show ws the stress state at the uultimate failu ure load (to the t left) and the stress state before the final failure (to thhe right), all stresses are given in MP Pa. The cracks that havee developed at a the two stages cann also be seeen in the figu ure. The cracck pattern fro om the FE-an nalysis is very ry similar to the t cracks in the tessted elementts. The failurre load is alsso in rather similar, s 104--123 kN for three of the keys and 82 kN foor the last onne. For this latter test wee suspect thaat a movemen nt might havve occurred in the rig, resulting in a horizonntal tensional frictional foorce in the sh hear key that lower the caapacity signifficantly.
hear keys. Figure 24: Concretee stresses and crack patttern for unrreinforced sh
2.2.2 L Large sca ale labora atory tests s The largee scale tests are a described d more detailled in chapteer 5. In this section s we juust focus on the t results concerninng the shear keys. The testeed key was in i this case the t smaller oone, with thee smallest width of 400 mm. This keey always occurs inn a couple wiith at centre distance d of 9980 mm. Thee larger shearr keys that haave been testted have a width of 500 mm, seee section 2.2.1. The propportion of sheear reinforcement betweeen the smalleer key and the largerr key is 5/7. During alll tests, with varying load ds and load ssituations (neegative/posittive bending moment) strrains were measuredd on two diffferent rebarrs in one of the shear keeys, the resu ult from thesse test is preesented in chapter 55. During thee last test, thee shear key w was loaded with w the max ximum load ffor the hydraaulic jack, 700 kN, sstill there weere no sign of o cracking. F FE-analysis shows s that ab bout 150 kN N should be trransferred
31
through tthe most loaaded shear keey in this sittuation. In orrder to get a shear key ffailure, a rigiid support was introoduced at thhe opposite side s of the j oint in order to transferr as much off the load ass possible through tthe joint, see Figure 25.
Figure 25: Load situ uation for th he final sheaar key test. d right between the two shhear keys thaat transfer The suppport was locaated 0,450 m away from tthe joint and the largest part of thee load. This location of the support should avoid the force ffrom just en ntering the support aas a compresssive strut thrrough the sheear key. Meaasurement shows that a foorce of 546 kN k will go done in tthe new suppport. The FE E-analysis givves a supporrt reaction off 557 kN, annd a shear keey load of 258 kN. Also underr this load, the t shear keeys remained d intact. Th he only failuure that occu urred was concrete crushing at the t support, due d to a veryy limited sup pport area (45 5x180 mm).
2.2.3 C Conclusio ons The prevviously used reinforced d point to point shearr keys (500 0 mm widee, 7-Ø12 mm m shear refinforceement) are fully fu capable of transferriing shear forrces > 300 kN. k For bridgges with a stteel girder spacing uup to 5,0 m, the traffic lo oads accordiing to Eurocode will nev ver exceed 2 16 kN (LM--1 = 1,5, 2x300 kN N axle load, lane l 1 9 kN/m m2 and lanex 22,5 kN/m2). The smalller shear keeys have a reinforcemen r nt ratio of 5//7 compared d to the largeer keys. Thee ultimate capacity should at leeast be > 214 4 kN. For brridges with a steel girdeer spacing upp to 5,0 m, the t traffic loads acccording to EN N 1991-2 willl never exceeed 130 kN in n this case. Concerniing the desiggn model, wee recommen d a simple and a safe apprroach using the well kno own shear resistancee formulas, from Euroccode, for a section with h inclined shear s reinforrcement. Wh hen shear reinforcement are useed locally, with inclined rrebars in onee line then the following eequation can n be used,
VRd ,s Asw f yw sin where Asw fyw α
( 13)
aarea of the shhear reinforceement yyield strengthh of the shearr reinforcem ment aangle of the shear s crack ( 45° observeed in the testt) innclination off the shear reeinforcementt
A previoously perform med fatigue test t indicatess good fatigu ue resistancee for shear kkeys in open ning joints (Hällmarrk et al, 20099). A failuree in the conccrete coverin ng is the mosst probably ffailure if thee covering layer is tooo thick. Aftter the test we w have founnd out that it is not a good d idea to deccrease the deepth of the concrete shear key. The T reinforceement bars inn the male-feemale connecction should overlap each h other, to make surre that the shhear keys con ntinues to traansferring fo orces over thee joint, evenn if there is a failure in the concrrete cover. This T is an issue concernning the robu ustness of th he constructioon, to ensurre that the shear keyys have a suffficient post failure f capaccity. The scattter in the tesst results mak kes it hard too propose leess shear rein nforcement iin the shear keys. The cost beneefit using Ø88 mm rebars instead of Ø Ø12 mm (34 rebars per element) e is aalso quite sm mall. There are possibbilities to im mprove the reeinforcementt layout in the shear keys. The layout studied in th his project has beenn proved to be b good eno ough, but thee work contiinues lookin ng at the posssibilities to make the design off the shear reeinforcementt even better.
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b taken into o account, whhen designin ng the reinforrcement, is tthe tolerancees that can One thingg that shall be be achievved. In this project we have experrienced both good and bad b workmaanship at pro ofessional concrete workshops. In Figure 26 6 below, the rreinforcemen nt in the firstt element forr the shear key k tests is shown (too the left) toogether with the reinforceement in a sh hear key for the large scaale tests (to the right). It is quitee obvious thhat the precission of the rreinforcemen nt work variees quite a loot, even if th he work is done by professionall at concretee workshopss. Bridge deesigners havee to take thhis into acco ount when choosingg type of rebaars. The SX-rebars in thee shear key to o the left in Figure F 26, w was for examp ple harder to get in tthe right possition comparred to the EX X-rebars to th he right.
Figure 26: Shear keyy reinforcem ment with baad accuracy y to the left, good to the right.
2.3 A Alternativ ve Design n with linear shea ar key Within thhe scope off the projectt, a bridge w with integraal abutments and an innnovative prefabricated bridge-deeck has beenn designed. The T bridge iss to be built in the South h of Germanyy close to th he town of Greißelbaach. It is currrently runniing through the approvall process by national andd regional au uthorities. The bridgge should bee completed in 2012. Thee bridge is designed d with h a span of 333.5 m and a width of 16m in oorder to carryy three lanes of traffic, a bicycle lan ne and two pedestrian ssidewalks cro ossing the ST 2200.. Main innovation of th his bridge iss the accelerration of thee constructioon progress by using prefabricated deck-ellements with dry joints am mong each other o that aree not connectted to the lon ngitudinal bearing ssystem. Due to the to thee new constrruction meth hod, particular requiremeents were im mposed by both locaal and nationaal authoritiess. The briddge consists of two (maiin) compositte-girders with w a steel-b box girder an and a concreete-slab in longitudiinal directionn. The constrruction heighht of the maain girders iss 95cm in mi mid-span and 245cm at the abutm ments with a constant height h of thee concrete sllab of 20cm. The widthh of this con ncrete-slab varies froom 140cm inn mid-span to o 385cm at tthe abutmentt. The Concrrete used is a high-quality y C 70/85 for the cooncrete-slab. In transvversal direction 16m long g and 2,50m m wide prefaabricated, pree-stressed cooncrete deck k elements are used.. The heightt of the deck k-elements vvaries betweeen 35cm in mid-span too almost 65cm at the cantilever arms. Thesse transversaal deck-elem ments are pree-stressed by tendons ø122/4 in its lon ngitudinal direction which indeeed is bridge transversal ddirection. In bridge long gitudinal direection, 11 jaccket tubes ø90mm aare assembleed to insert stressing-cab s bles in order to pre-stresss the deck-ellements to eaach other. The conccrete used forr transversal elements poossess the hig gh compressiion strength oof C 70/85. There is no shear-connnection bettween the traansversal eleements and th he longitudinnal bearing system so that only vertical forcces can be trransmitted. T Thus, there iss no compossite action beetween concrrete deckelements (transversaal bearing system) s andd the main girders (lon ngitudinal bbearing system). The transversal elements are a connecteed to each othher by a dry key and slot joint which are being prre-stressed in longituudinal directtion by cablees in jacket ttubes. This construction c method is seet up with th he specific aim to aaccelerate thee constructio on process eespecially fo or large brid dges with siignificant am mounts of concrete--slabs. As noo in situ concrete is needded, great savings of con nstruction tim me are suppo osed to be aspired. T The bridge iss designed in n accordancee to the Germ man DIN-FB 104 which iis based upon n DIN EN 1994-2:2006-06 and 1994-2-1:1994-02.
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Mid-Span Figure 27: Cross-Section of the Bridge in M
dinal section n of the bridgge Figure 28: Longitud ouple the lon ngitudinal annd transversaal bearing The consstruction metthod is baseed upon the iidea to unco systems aand thereby to t acceleratee the construcction progress. Load tran nsmission in longitudinall direction of the briidge is accom mplished by the longitudiinal girders only o while in n transversall direction it is assured by the prrefabricated deck-elemen nts. For reas ons of consttruction prog gress, the briide deck con nsists of a directly ddriven on prrefabricated concrete slabb which sho ows already the desired bbridge surfaace profile includingg the kerbstones. In orderr to provide a sufficient durability d of the directly accessed briidge deck, a very deense, indefiniite rough and d crack-free cconcrete is chosen. The transsversal deck elements aree assembled on the longiitudinal girdeers on speciaal bearing strrips which provide uunobstructedd horizontal movementss and deflecctions of thee transversall deck elem ments. For reasons oof accuracy, the transverrsal elementss are also grrinded in this contact-zonne to the lon ngitudinal main-girdders. In longituudinal directtion of the brridge, the traansversal elements possess dry jointss (principle of o key and slot). Forr shear forcees the joints are a designedd in accordan nce to a strutt-and-tie moddel. The con ntact areas are grindded exactly too the desired d geometry aat the fabricaation-plant so o that a fast construction n progress on site iss possible. In I longitudin nal directionn of the bridge, the deck k-elements aare pre-stresssed by 11 cables assembled in jacket tubes ø90mm ø in a pattern of 1,,48m. Calcullation of the pre-stressing g forces is crucial foor the deck-ddesign.
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Figure 29: Principlee of the Key and Slot Joiint
Figure F 30: Key K and Slott Joint (Con ncreting Testt)
gned to assurre a constantt compressio on in the areaa of key and d slot joint The pre-sstressing forcces are desig under eveery load com mbination to prevent wateer to intrudee into the gap p. Further onn, a jolting fo or passing cars can be excludedd by this proccedure as weell. In the top pside contact-zone, a seaaling establisshed from segmentaal-tunnelling is assembleed additionaally in orderr to preventt water and aggressive media to penetratee the gap andd thus to asssure a high qquality and great g durabillity of the coonnection. In n order to provide tthe possibilitty to change single deckk-elements, th he tendons in n the jacket tubes are no ot injected with morrtar and thus function as “external” “ teendons by staatic means. As the grrinding-proceess is very prrecise and acccurate, the gap g between single transvversal elemen nts can be considereed to be fabrricated exactly as provideed in the draawings. The tolerance forr the grindin ng process can be liimited to leess than 1/10 0mm. Summ marizing, it can be stateed that the ““open” gapss between transversal deck-elem ments in brid dge longituddinal directio on can be handled h by ppre-stressing forces, a sufficientt sealing andd a very high accuracy forr producing and a grinding g the gap surffaces.
Figure 31: Deck-Eleement beforee Grinding
Figure F 32: After A Grindiing the key-sslot joint
To secure the superstructure in its position, ttwo construcctional elements are neeeded. The deeck slab is prons behind d the abutmeents. These aprons a are supportedd in longituddinal and traansversal direection by ap fixed elasstically betw ween the abu utments by m means of perm manently elaastic material al and absorb b breaking forces as well. In trannsversal direection, the poosition is asssured by concrete cleats oon the elemeents at the inside of the girders, in case the pre-stress p is innterrupted fo or example when w replacinng elements.
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3 DURABILITY SLAB-SLAB For a multi-span composite bridge having prefabricated deck elements, that are not connected to each other by other means than offered by match casting, the transverse joints will and must open over intermediate supports as the bridge is affected by traffic loading. This is quite all right from a statical point of view, but may cause damage to the water insulation layer and the asphalt wearing layer. Large longitudinal strains will occur in these layers if the insulation and the asphalt are put in place without any special attention to these small but critical regions. The results of a series of fatigue experiments addressing this problem are reported in this chapter. Each test specimen simulates a part of two real deck elements – i.e. a 66 cm wide slice cut out in the longitudinal bridge direction – such that the joint between them can be tested. The two bridge deck slices were loaded simulating a gap opening of a magnitude that can actually occur over an intermediate support due to traffic loading and other loading conditions such as temperature and concrete shrinkage. It is the fatigue resistance of the surface layer with respect to water penetration and de-bonding that have been tested. The insulation and asphalt layers are attached to the top of the concrete deck elements. The specimen was tested by pulling the two deck elements apart then pressing them back and so on until fatigue failure of the insulation layer was achieved. The entire experiment was carried out under wet conditions such that the asphalt surface always had a 1 cm pond of water mixed with antifreeze on top. Failure was assumed to have taken place when the water penetrated into the underlying concrete. Temperature is crucial for bitumen based product, thus the whole experiment was carried at a temperature of -20°C as soon as temperature equilibrium of the climate chamber was reached. This means that the experiments were carried out under fairly extreme conditions. The experiments were run under displacement control such that a well-defined deck-to-deck opening was achieved during each load cycle. Each test was set of at a deck opening of 0,4 or 0,5 mm which were then ramped up to a bit more than 2 mm, unless failure occurred before. In total, 6 specimens were tested, with 4 different arrangements of the insulation and asphalt layers in the neighbourhood of the deck to deck joint. In particular the first 4 were all slightly different, while the last two had the same properties as the 4th specimen. The first specimen had the insulation and asphalt arranged as if there was no joint to worry about, while all other specimens were modified such that their fatigue properties should be improved. In short the first specimen sustained 159 000 load cycles, the second 241 000 cycles, the third 3,05 million cycles. Specimens 4 to 6 survived from 4,37 to 5,93 million load cycles.
3.1 Test setup General aspects of the test specimens are given in this section. The test rig, loading and measuring equipment together with the climate control facility is also described.
3.1.1 Test specimens Here general properties of all 6 specimens are given. The exact detailing of the insulation layer over the joint is described in Sections 3.2.1 to 3.2.4. Figure 33 shows the principal configuration and dimensions of the test specimen. The two bridge deck elements were simulated by two concrete blocks, 1 820 mm in length and 240 mm in depth. Their width in the transverse bridge direction was 660 mm. The complete specimen can be viewed as a slice of width 660 mm cut out from two deck elements in the longitudinal direction of the bridge. The two concrete blocks were heavily reinforced in order to prevent premature failure in the concrete, but also to make a safe and well defined load transfer into the concrete possible, as indicated in Figure
37
ment 12 fullly threaded M20 bars of o grade 8.88 without an ny surface 34 a. Ass longitudinaal reinforcem treatmentt was used in each con ncrete blockk. The ends of these baars extendingg into open air were connectedd to the loadding equipmeent as shownn in Figure 34 4 b. Eleven ordinary o recta tangular reinfforcement stirrups ssurrounded thhe longitudin nal reinforcem ment in each h concrete block. The nosse of each bllock at the deck-to-ddeck joint (or shelf) was reinforced bby 8 + 7 U-sshaped reinfo orcement stirrrups and 2 transverse t bars at thhe very tip. All A traditionaal reinforcem ment bars werre of grade B 500 and haad a diameterr of 8 mm. In additioon, 4 smoothh circular ro ods were mouunted at mid d-thickness going g right tthrough each h concrete block at supports BC C, D, GH and d loading po int F, see Figure 33. Theeir ends outsside the conccrete were used to ttie down thee concrete bllocks, i.e. prreventing uplift in case of o negative ssupport reacctions, see Figure 366. At point F the bar wass also used too apply the simulated s ax xle load. Insiide the concrrete but as close as ppossible to thhe side surfacce these barss were surrou unded by two o extra U-shaaped stirrupss.
Figure 33:
a) Figure 34:
Dimensions of th he test specim men. Its wid dth perpend dicular to thhe plane of the t paper is 6600 mm. The principal ssupporting and loading arrangem ment are alsso shown schem matically.
b) a) Reiinforcementt on both sid des of the joint. b) Load d transfer innto the specimen.
The cylinnder-to-cube strength waas 45/55 MPaa for test speecimens 1 to 3 and 50/600 MPa for speecimens 4 to 6. Thee strength class was incrreased due too problems with w the frettting resistannce of the sh helf in the deck-to-ddeck joint. The T material properties p off the concrette met the requirements aaccording to EN 1992 (Eurocodde 2). The sttrength classes used are likely to maatch those ussed for real bbridge deck elements. The maxiimum diameeter of the ag ggregate was 16 mm and the t consisten ncy grade waas S4. Slurry y additives and an aiir entraining agent were also a used givving an air co ontent of at leeast 4,5 %. men were maatch cast in the sense thaat the two haalves were The two elements maaking up onee test specim m thick steel sheet bent iinto the shap pe of the deeck-to-deck jjoint. This steel s sheet separatedd by a 1 mm separatedd the formwoork into two halves. h
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The concrete was properly vibrated by means of two hand held poker vibrators (one 28 mm and one 45 mm) during casting. After about an hour the top surface was levelled flush with the top edge of the formwork. The top surface was rubbed down after 3 to 4 hours using a hand float. Finally it was finished by steel trowelling within 4,5 to 5,5 hours after casting. After casting the concrete was left to cure for at least 28 days, whereof one week in the plywood formwork with the top surface covered by a plastic film. Then the formwork was removed and the specimen left to cure in the dry conditions of the laboratory. Shortly before the concrete blocks were put in the rig they were brought outside and the top surface was sand blasted where after the surface was cleaned from dust by blowing with compressed air. This treatment was performed according to specifications of the Swedish road administration. The surface was then protected by a clean plastic film until the surface primer could be applied. For most specimens the primer was applied before the specimen was put in the rig and fixed to the loading equipment. The primer, which is approved by the Swedish road administration, was a two component primer based on methyl methacrylate with product name “Beta A-primer” from DAB Domiflex AB. First, a base layer was rolled onto the surface and when still wet quartz sand (grain size 0,7 mm) was sprinkled onto the surface. After curing the excessive sand was brushed away and a second layer of primer was applied. The insulation layer/layers and the layers of mastic asphalt were applied after the concrete decks had been properly installed in the rig. The two slabs were gently pushed towards each other before final fixing to the loading arrangement. The water proofing membrane was made of a 5 mm thick bitumen impregnated polyester/glass-fibre felt, whereof the downward surface was made of 3 mm weldable bitumen. The bitumen used for the skeleton was polymer modified. The product used was “BETA 6000 SA” from Nordic Water Proofing AB. This product has good low temperature properties, with regard to strength and elasticity and is commonly used for water proofing of bridge decks in Sweden and is of cause approved by the Swedish road administration. The exact layup around the deck joint is given in Sections 3.2.1 to 3.2.4 as the test of each specimen is described. But basically the membrane was welded to the primed deck surface without the edge railings of steel mounted. Then all the railing plates were screwed onto all sides of the concrete, while the membrane was folded up 90 º along all edges. Finally, the membrane was welded to the railings and arranged in the corners such that a water tight basin was formed. The membrane was not welded to the steel railings along 30 cm on each side of the joint, which was done to avoid failure of the membrane above the concrete surface. After this two layers of 3 cm mastic asphalt was applied. The second layer was applied 30 to 60 minutes after the first layer. Both layers were taken from the same batch. The total thickness was targeted at 6 cm in the region above the deck joint, which in reality varied between 5,5 cm to 6,5 cm. The mastic asphalt had a maximum aggregate size of 11 mm and was based on polymer modified bitumen. The bitumen content was 7,9 % and the amount of filler material was 28 %. It was delivered by NCC Binab AB and specified as type PGJA 11 and produced according to recipe number 16. After this the specimen was left to settle for at least 2 days. Meanwhile, the measuring equipment was attached and the walls and roof of the climate chamber (freezer) was put in place. Before closing the room the surface basin of the specimen was filled to about 1 cm with a mixture of 50 % water and 50 % antifreeze (propylene glycol). Finally, the load actuator and the freezer unit were started and everything checked before the room was sealed.
3.1.2 Test rig A schematic sketch shows the test rig from above in Figure 35. But note that the number of tension/compression bars was reduced to 4 in the real rig and that each group of bars was attached to a group of 6 M20 reinforcing rods in the concrete as indicated in Figure 34 b. In reality all transverse frame members were made of two IPE 400 profiles with their web panels facing each other. Each longitudinal frame member was in the same way made of two IPE 220 profiles. The frame was centred with respect to the mid thickness of the concrete and the mid longitudinal line of the specimen. Each group of 6 M20 rods were arranged in the concrete such that 3 of them had their centre
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t top face and 3 of theem 50 mm from fr the botttom face, givving a 140 mm m centre placed 500 mm from the distance in the vertical direction. Their spacinng in the traansverse direction was 800 mm and thee smallest centre disstance to thee longitudinaal concrete eddge was 40 mm. m Most off these detaills are shown in Figure 34.
Figure 35:
Princcipal sketch of the test riig as seen frrom above.
wo longitudin nal supportinng beams maade of IPE The entirre frame inclluding the test specimen rested on tw 400 profi files of lengtth 7,0 m. Theese longitudiinal beams was w connecteed to the rigg and the speecimen in such a way that both positive and d negative suupport reactio ons could bee counteracteed by them at supports BC, D, G GH and loadiing point F as a indicated iin Figure 33. The test speecimen was ffixed to the frame f and the framee to the longgitudinal beaams at the lleft end of Figure F 33. The rest of thhe rig and all a vertical supports of the specim men were flo oating on paiirs of plastic pads. Each pair p had onee pad attached towards the longiitudinal beams and onee pad towarrds the speccimen or thee rig. All ppads were reectangular 120×80×20 mm blockks of PTFE (polytetrafluuoroethylenee) having a very v low friiction coefficcient. The loading eend of the rig r was also prevented ffrom moving g transverselly, in relatioon to the lon ngitudinal beams, byy another sett of 4 pairs of o PTFE-bloccks. Finally each e verticall support of tthe specimen n was also strapped down to the longitudinall beams in orrder to preveent uplifting at the suppoorts. These sttraps were only prevventing uplifft and did no ot restrain diisplacements in the longiitudinal direcction. Some strapping details caan be viewedd in the phottos of Figuree 36. The strraps were acctually only nneeded as lo ong as the mastic assphalt had noot cracked ov ver the deck--to-deck join nt. Then largee uplift forcees, which cou uld not be counteraccted by the self-weight of the concrette and the sim mulated axle load, were ppresent at thee supports due to thee eccentric looading with respect r to thee centre of th he asphalt lay yer. The actuaator used waas of model MTS M with a capacity of ±350 kN. It is showed inn Figure 37 a together with the ccomplete tessting rig and a specimen iin place.
3.1.3 M Measuring g equipm ment The totall load was measured m with h an MTS loaad cell moun nted on the actuator. a Its aaccuracy is better b than ±1 kN. A Also, the dispplacement delivered by thhe actuator was w measured d by its interrnal transduccer, which has a preccession betteer than ±0.05 5 mm. Displacem ments were also recorded at the deeck-to-deck joint. Here two linear vvariable disp placement transduceers (LVDT) were used. Their T total raange and acccuracy were 10 mm ±0.055 mm. One transducer t was mouunted on eachh longitudinaal edge of thee specimen and a in level with w the bituumen membrrane. They were not directly attaached to the concrete butt to the steel railings, wh hich in turn w was securely y bolted to the concrrete by fully threaded M8 8 rods goingg transversally through th he concrete sppecimen. Six x M8 rods were usedd for each cooncrete block k.
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a two locatioons. One sen nsor was han nging in the air about half a meter The tempperature wass measured at from the jjoint while the t other sensor was fitteed into a small hole drilled in the asphhalt 2 to 3 dm m from the joint. b a video camera c whichh could captture video Leakage was monitorred through visual inspeection aided by ken every 5th minute duriing the tests. By this it or take snnap shots at regular interrvals. Snapshhots were tak was possible to pinpooint the onset of leakage down to a precision p of about a 300 loaad cycles. Th he camera was placed such thatt the photos showed the area under the t joint, wh here water drrops fell ontto a white collectingg surface. If the leakage appeared diirectly over the t joint the water fell sttraight down n onto this surface. IIf the water came out on n the longituudinal concreete edges aw way from thee joint it wass led back onto the collecting suurface throug gh a system of thin foldeed metal sheets acting ass gutters. On ne of these photos iss shown in Figure F 36 b. Here H no leakkage is visib ble but all thee dust and sm mall grains created c as the concrrete parts grind each otheer at the joint nt is clearly seen. The resolution of thhe photo wass such that a single ddrop of water could be sp potted on a cclean collectiion surface, while w a few drops are neeeded on a heavily ddusted surface.
a) Figure 36:
b) n straps on the rig. b) Leakage a) Leeakage detecction camerra and two hold down detecttion photo. Test T 4 after more than 5 million cycles.
b)
a) Figure 37:
a) Th he complete test rig wiith a specim men in placee. b) The riig inside the climate cham mber with thee roof lifted off.
3.1.4 C Climate control The entirre test rig waas contained within a dem mountable co ontainer with h an inner areea of 1,9×7,2 2 m2. Each wall wass one unit annd the flat ro oof was madde of two un nits. When changing a teest specimen n one roof
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element and the two longitudinal wall elements were lifted away. The wall and roof elements were traditional framed elements made up of 45×95 mm2 timber studs with OSB-sheeting on both sides. Plastic film and weather strips of rubber were used to prevent air and moisture from penetrating into the chamber. Traditional mineral wool of 95 mm thickness was used as heat insulation between the studs, which were spaced at 600 mm. The floor was made of 22 mm particle board floating on top of a 100 mm layer of expanded polystyrene. As a protection, a rubber sheeting of 1 mm thickness was placed on top of the particle board. The long steel beams of the rig just rested on top of this rubber sheet. One of the long wall elements also had a small door, making regular inspections possible. The freezer unit (Midifrost-30-V from Frigadon AB) was placed towards the roof in the gable wall closest to the test specimen. This unit had a capacity of cooling a, relatively well insulated, volume of 35 m2 down to about -25°C. When everything was in heat equilibrium the temperature in the asphalt varied less than 0,5°C and the air temperature less than 4°C. All climate controlling parts, i.e. the chamber and the freezer unit, are shown in in Figure 37 b.
3.2 Test procedure and results Here individual properties of each test specimen are given together with the test results. In total 6 fatigue tests were performed. All 6 specimens looked basically the same except for the arrangement of the water proofing membrane right above the deck-to-deck joint. In all cases the load frequency was targeted at 1 Hz, but was usually slightly less (down to 0,95 Hz) depending on the load level and applied displacement. Note also the simulated axle load which was applied by means of the cantilever system shown in Figure 33. The total load at F was 20 kN and it was maintained long after the mastic asphalt had cracked, but most of it (90 %) was removed after the joint displacement had passed 1,60 mm. This was done to reduce the abrasion of the concrete shelf. All tests were started at room temperature. First the joint opening was adjusted to 0,35 mm, i.e. the set point that is the mean value of the cyclic starting displacement to be applied. Then the amplitude (or span) was increased to 0,50 mm. These are values that were used for specimens 1, 2 and 3. For specimens 4, 5 and 6 the corresponding values were 0,30 mm and 0,40 mm, respectively. Then all equipment was checked and the testing chamber was closed. It generally took about 2 days to reach air temperature equilibrium in the chamber and one more day to reach temperature equilibrium within the asphalt. During this period both the set point and span had to be adjusted on a regular basis in order to maintain the desired displacement variation. In general both the set point and the span decreased due much faster cooling of the steel rig than the concrete.
3.2.1 Test 1 Here only one layer of membrane was welded to the entire concrete surface and the mastic asphalt was poured onto the entire membrane surface. This resembles a joint for which no extra detailing around the joint is made, i.e. the joint is treated as any other surface of the bridge deck. The mastic asphalt cracked right through after less than one day (86 000 load cycles) at an asphalt temperature of only -5°C, but no leakage was observed. Now less force was needed to pull apart the deck elements and consequently the amplitude of the joint displacement was increased to a bit more than 1 mm because of the elastic properties of the steel rig. The problem was detected within an hour and the span was reduced to 0,6 mm. This level was maintained up to 159 000 cycles at which leaking water drops were detected. The temperature was now -16°C in the bitumen and the leakage gradually became worse. The specimen was inspected after 168 000 cycles. It was observed that leakage was a fact on both longitudinal sides of the specimen, where after the first test was shut down. After dismantling the specimen it was observed that the crack in the mastic asphalt was perpendicular to the longitudinal direction and located about 4 cm aside from the joint opening. The membrane had either de-bonded from the concrete or pulled off large pieces of the concrete within this these 4 cm. The
42
d at both coorners where the membrane was foldded up on the t railing leakage aappeared to have started plates.
3.2.2 T Test 2 Test 2 waas similar too Test 1 exceept that a de--bonded region with a total width of 28 cm (14 cm m on each side of thhe joint openning) was created betweeen the memb brane and th he concrete. T This was ach hieved by placing a 28 cm wide strip of undeerlay felt (YA AM 2000 fro om Icopal AB B) over the jjoint. This test went on as planned p up to o 241 000 loaad cycles. Bu ut the displaccement ampllitude had deecreased a bit more than expectted during th he night. It w was, therefo ore, decided to make an inspection within w the climate cchamber. No problem waas encountereed, so while still in the ch hamber the sspan and set point was adjusted to the desireed amplitudee of 0,5 mm.. During thiss process thee mastic asphhalt suddenly cracked right throough with thhe crack peneetrating throough the mem mbrane as well w causing ssevere and immediate leakage. T The crack was w perpendiccular to the loongitudinal direction d and d located righht above the joint gap, deviatingg from the joiint by no mo ore than 1 cm m. When thee mastic aspphalt suddenlly cracks thee joint displaacement willl, during thee on-going lo oad cycle, increase tto more thann 1 mm due to the elasticiity of the steeel rig. Since the asphalt iis kind of glued to the membranne very largee and local straining s willl occur in th he membrane at the veryy crack tip coming c in from the asphalt.
3.2.3 T Test 3 The clearr lesson from m Tests 1 and 2 is thatt neither the asphalt nor the concrette can be fix xed to the membranne in the regiion surrounding the deckk-to-deck join nt if it shall be b able to susstain an open ning of up to 2 mm ffor a reasonaable number of cycles. Inn Test 3 thiss was achieveed by creatinng a de-bond ded region between all membranne layers, beetween concrrete and mem mbrane, and asphalt andd membrane. Each debonded rregion was a 30 cm wid de strip (15 cm on each h side of the joint), exccept between the top membranne and the assphalt where the strip waas only 20 cm m in width. Each E strip connsisted of tw wo 0,1 mm thick PTF FE-foils staccked on top of each otheer. PTFE can n sustain a temperature oof 300°C forr a longer time but melts aroundd 350°C. Ev ven though thhe temperatu ure of the torrch flame is uup to 700°C it proved possible to weld thee membrane all the wayy up to the edge of the PTFE-film m. This wass done by protectingg the PTFE--film with the membranee itself, i.e. sttarting the welding w at thee joint and rolling r out the membbrane in botth directions. For this sppecimen, step ps in the con ncrete was caarved out towards the joint suchh that two short s membrrane pieces ccould be fittted underneaath the top m membrane which w then became aalmost perfecctly flat. Thee layup is illuustrated in Fiigure 38 a. Note N that the bbottom mem mbrane has no real fuunction but was w needed as a the steps inn the concrette was given the wrong shhape during casting.
a) Figure 38:
b) Arran ngement of water prooofing membrranes and de-bonding d sstrips over the t deckto-decck joint in a) a Test 3 and d b) Tests 4, 5 and 6.
00 load cyclees at an asph halt temperaature of -9°C C, but no leaakage was The asphhalt cracked after 119 00 caused byy this. The suddenly s incrreased displaacement span n was reduceed down to 00,6 mm and repeatedly r adjusted to this level as the tempeerature droppped down to the equilibriu um value off -19°C (to -19,5°C ), a
43
temperature which was then maintained during the entire test. After temperature equilibrium had been obtained the displacement span was increased by 0,1 mm every day (except during weekends where no increase was made). This process was started at 593 000 cycles at which the span was adjusted to 0,7 mm. At precisely 2,00 million cycles the displacement span reached the target value of 2,0 mm. After this the span was held constant at 2,0 mm until failure (leakage), which was clearly detected at 3,05 million cycles. The leakage started as soon as some small crack had penetrated the top membrane probably close to one of the transverse edges of the de-bonding strips. The crack was so small that it was impossible establish its position without doubt after the specimen had been taken apart. The second membrane layer did not crack. Instead the water penetrating the first layer found its way out through the voids that were unintentionally present between the first and second membranes along the fold lines towards the side railings. The welding of the top layer to the second layer was by mistake not properly done up onto the side railings. Hence the water could find its way out without passing through the second membrane.
3.2.4 Tests 4, 5 and 6 All these tests were carried out in the same way. The temperature equilibrium was -20°C (to -20,5°C ) for all of them. The concrete surface was flat without any steps. Two membrane layers were used having de-bonding strips arranged as given in Figure 38 b. But this time the two membranes were properly welded to each other on the vertical railing sides, i.e. from the de-bonding strip between them and away from the deck-to-deck joint. For these tests the joint displacement was ramped up to the target value of 2,0 mm at a faster rate compared to Test 3. The joint displacement was adjusted to 0,8 mm after through cracking of the asphalt. Where after the displacement span was increased with 0,2 mm per day (except on weekends) until the target was reached. Asphalt cracking occurred after 137 000 cycles for Test 4, 158 000 cycles for Test 5 and 144 000 cycles for Test 6. The first increase to a 1,0 mm displacement was, after reaching temperature equilibrium, done at 423 000 cycles for Test 4, 573 000 cycles for Test 5 and 467 000 cycles for Test 6. For Test 4 no leakage had occurred when 5,00 million cycles was passed. It was then decided to increase the displacement span in intervals of 0,2 mm. A span of 2,2 mm was held until 5,23 million cycles, 2,4 mm until 5,41 million cycles and 2,6 mm until 5,93 million cycles at which leakage finally occurred. The same procedure was followed for Test 5, which failed at 5,33 million cycles when subjected to a displacement span of 2,4 mm. Test 6, however, failed at 4,37 million cycles at the target displacement of 2,0 mm. Tests 4, 5 and 6 all failed by through cracking of the membrane close to the transverse edges of the debonding strips. But the cracks through each membrane layer did in general not appear over each other. This proves that the two different widths of the de-bonding strips worked as intended, i.e. when the first crack formed in the top membrane it was stopped from penetrating into the second membrane by the debonding strip in between them. And the bottom membrane could now work as a second protection barrier. Also, it is natural that the first crack came in the top membrane due to greater straining because of the smaller width of the de-bonding strip in between the asphalt and the top membrane.
3.3 Conclusions and discussion From Tests 1 and 2 it is concluded that it is impossible to allow any substantial deck-to-deck opening if one or more water proofing membranes are just welded to the deck and then paved with asphalt. This is also in line with the tests performed by Edwards and Westergren (2001). It will not matter if one, two or even three membranes are used if they are fully welded to each other. A crack will quite naturally penetrate through all of them. From beneath the deck-to-deck opening acts as a sharp crack tip and from above any cracking through the bottom face of the asphalt will also act as a sharp crack tip. As a worst scenario we will have a crack tip on each side of the membrane. For severely low temperature cases it is
44
unlikely that a bitumen based membrane can sustain displacement fluctuations of more than 0,2 mm for millions of cycles. From Tests 1 and 2 it is also clear that the elongation of the membrane must be distributed over a fairly large length in order to reduce the strain to acceptable levels. From Tests 3 to 6 the following conclusions can be drawn:
If just one single water proofing membrane is used some kind of artificial de-bonding between the membrane and concrete as well as between the membrane and asphalt must be present to substantially improve the fatigue resistance. The width of such regions must be about 2 dm. For this solution to work it is probably also necessary to artificially control the cracking of the asphalt layer such that it is more or less located over the deck joint opening rather than over the edge of the de-bonded region. Otherwise, the de-bonding strip cannot prevent the crack from growing into the membrane.
If two or more water proofing membranes are used the above statement holds but the de-bonded regions must increase in width when going from the top towards the concrete surface. Otherwise, the de-bonded region will not be able to stop a crack in an upper layer from growing into an underlying one. This reasoning holds as long as the cracks are formed and propagates from the top and downwards, which is the most likely scenario over an intermediate support of a bridge deck. The width increase should be of 1 dm in magnitude and definitely not less than 5 cm, even if these regions can be placed with great accuracy.
It is also clear from these tests that natural de-bonding between the material layers, even between the concrete and the bottom membrane, will not occur before the layer of mastic asphalt cracks right through over the deck-to-deck joint. Using PTFE-film as a de-bonding layer is of cause not very suitable from a production point of view. But was in this study just used to demonstrate how the fatigue resistance can be improved by working with de-bonded regions. It would be possible to develop water proofing membranes having these properties built in, herby reducing the number of required work hours to just a little more than for traditional water proofing and at the same time minimising the risk for assembly faults. Why was the deck-to-deck opening displacement chosen to have a target value of 2,0 mm in this investigation? The reason is that the opening is mainly controlled by the straining of the top flange in the undelaying steel girder if full composite action is assumed. For a steel flange with fy = 355 MPa and a spacing of 1,8 m between deck joints the deck-to-deck opening would, by using Hook’s law, be 3,0 mm before yielding of the flange. In reality the value must be less because far from all load is variable load. The curvature of the steel girder will, on the other hand, increase the value in relation to the thickness of the concrete. Also the shrinkage of the concrete will increase the value, but this effect is static and most shrinkage took place before assembly. Finally, partial composite action and shear lag effects may also influence the value to a minor extent. Hence, 2,0 mm seamed a suitable value. There are two drawbacks with the solution according to Tests 4 to 6 that must be considered. The first is that this deliberate de-bonding is usually not allowed in present regulations, because it is suspected that the de-bonded region may grow when for instance passed by heavy vehicles. Here, the solution is to convince road authorities the small de-bonded regions can be accepted, especially if a slightly thicker paving is used on top of the membranes. The second is that water, dust and all kinds of pollutions will penetrate down to the membrane as soon as the asphalt cracks. This problem can be counteracted by using a rubber based dust and leakage barrier embedded in the asphalt above the joint (see Figure 39 ) as described in section 3.2.3 (Figure 40) of the Design Guide.
45
n of dust Figure 39: Use of a rubber cealiing to avoid penetration
46
4 SH HEAR CON NNECTION N GIRDER R –SLAB To gain composite action, a the giirder and thee slab elemeents have to be connecteed by meanss of shear connectors. Within the t project, two differennt means off shear conn nection weree under inveestigation. Clothoidaal shaped cooncrete dow wels were tessted in Wro oclaw Univerrsity. Clothooidal shaped d concrete dowels ooffer the possibility of lower toleraances for th he concrete elements annd the placin ng of the transversal rebars. Inn addition, the fatigue resistance is enhanced by the speccial clothoid dal shape. Furtherm more, a bridgee was design ned by SSF uusing a comp posite grillage for the asseembling process. This measure aims on an accelerated constructionn process by y using preffabricated m modules whicch can be assembleed on the connstruction sitee.
4.1 Laboratorry testing g on cloth hoidal sh haped co oncrete d dowels 4.1.1 IIntroductiion By RFSR R-CT-2006-000030 projecct compositee dowel connection has been establiished but on nly puzzle shape haave been tested in detaail, because no efficien nt technolog gy have beeen developed yet for fabricatioon of clothoidal shape th hat time. Beecause of hig gh fatigue resistance off clothoidal shape, s the searchingg for cutting technology was continuued and it fin nally succeed ded with devvelopment off modified clothoidaal (MCL) shhape cutting g line (Berth thelemy et. al. 2011). It I was deveeloped in Poland for realizatioon of innovattive railway bridge desiggned by Euro oprojekt Gdaańsk S.A. coompany (Buk kała et. al. 2011). Inn frame of KB B/117/14030 0/IT1-B/U/088 national Po olish research h some tests needed for realization r of the brridge were conducted. c On O the basiss of promisiing results of o first tests,, this new shape s was adopted ffor purposess of ELEM project. p As reealization off steel part of o this moderrn shear con nnection is fully autoomatic, folloowing the id dea to increaase degree off automation n and prefabbrication of composite c superstruucture of thee bridge, tessting program m was desig gned in fram me of ELEM M project to o combine conventioonal studs with w composiite dowels, oor even to substitute s thee studs by ddowels if posssible and economiccally justifieed. Accordin ng to results of tests reaalized in fram me of RFSR R-CT-2006-0 00030 and KB/117/114030/IT1-B B/U/08 projeccts, due to ddifferent stifffness of com mposite doweels versus stu uds it was possible tto superposee it with stud ds for ULS oonly, but for FLS and SL LS checks suuch an approach is not justified. According to t this and on n the basis oof analytical studies undeertaken, it waas decided to o test only gn segmentaal bridges wiithout any a new forrm of compoosite dowelss to define itts resistance, and to desig studs. Onn the other hand, h for desiign checks aat ULS it is possible p to combine it wiith conventio onal studs by simplee superpositiion according g to standardd formulas iff needed. The beariing capacity of a compossite dowel iss limited by steel s or conccrete failure. In a good deesign both failures oof a compositte dowel are balanced upp to the maxim mum load. Steel failuure, see Figuure 40, is lim mited: -
inn the ultimatte limit state by the shearr resistance in nfluenced by y bending, inn the serviceeability limit state by steeel yielding due d to bendin ng of the connnector comb bined with itts shearing and a local presssure to steell surface, inn the fatiguee limit state by b fatigue craacks due to cyclic c loading g.
a)
b)
c)
Figure 40: Failure modes m for steeel (draw. P Preco-beam project p (Preecobeam))
47
Concrete failure is characterized by several failure modes. Which mode finally occurs depends on the boundary conditions like geometry, concrete grade, reinforcement design, adding of fibers etc. The most up-to-dated formulas for concrete failure are presented in (P804). In the ULS concrete failure may by govern by shear of concrete or by pry-out cone. In FLS it should be verified that both shear and pryout criteria are limited to appropriate values, which were assumed to be 0,7 of ULS resistance. Overall 18 POST specimens divided into 6 test series were tested. Every series were design to enable conclusion about influence of some factor on resistance and ductility of shear connection with composite dowel: -
-
POSTs A-C – influence of web thickness on steel resistance, POST D – influence of number of dowels on its resistance (2 dowels in series D versus 3 dowels in series A-C, direct comparison between series B and D, as these POSTs were carried out with web thickness of 20 mm), POST E-F – influence of number of dowel strips and its spacing (direct comparison between series D and E-F, as these POSTs were carried out with web thickness of 20 mm and 2 dowels in one stripe).
It was needed to tests resistance of connection by two strips because structural solutions for segmental bridges proved, that it can be reasonable solution for twin-girder bridges.
4.1.2 Steel failure criteria According to POSTs formula describing ultimate resistance were created as shown below:
Ppl, MCL AULT ex tw f y where: AULT shape coefficient to determine on basis of tests ex spacing of dowels along web tw web (dowel) thickness fy steel yield strength As web thicknesses which were tested are within 15 – 25 mm, it is important that the final formula will be valid only for webs up to 25 mm. Test evaluation acc. to EC4 was done for all POSTs and on its basis differences between every series were calculated. As a basic test POST A was chosen (tw15), as in this test steel failure occurred with the biggest safety margin to concrete failure. Results from test series B and C are related to these from A series. t w, B f y , B 20 403 PRd , B 457,2 1,14 1,28 (1,28 / 1,14 = 1,12) PRd , A 400,9 t w, A f y , A 15 421
tw,C f y ,C 25 397 PRd ,C 547,7 1,37 1,57 PRd , A 400,9 t w, A f y , A 15 421
(1,57 / 1,37 = 1,15)
Increase of connection resistance with increase of web thickness is 14 and 37 % for webs 20 and 25 mm in relation to web 15 mm, and taking into account percentage increase of web thickness combined with its yield strength it should be 28 and 57 % respectively. It can be stated that the thicker webs the bigger resistance occurs, but safety factor should be assume. For tw25 this factor is of 1,15, so determined AULT factor should be multiplied by 1 / 1,15 = 0,87. This is safely estimation, because for POST C concrete failure starts to govern total resistance (but there is still small difference between 1,12 and 1,15).
48
n now calcullate: On basis of POST serries A we can AUULT , real
Vtest 556 ,8 0,353 e x t w f y 0 ,25 0 ,015 4421000
o AULT facto or should be in conformiity with EC4 4 so reduction on factor of 0,9 0 should Characterristic value of be impleemented. Alsso factor 0,8 87 calculatedd above as a safe estim mation of weeb thickness influence needs to bbe taken intoo account. Th his way we ccan have: AULT , Rk 0,9 0,87 AULT , test 0,9 0,87 0,353 0,276 0,25 2
Finally, ttaking into account a that AULT value w was calculateed on basis of a few PO OST specimeens results and that in Germany coefficient of o 0,25 is asssumed for MCL M shape also a on the bbasis of latest research (P804), itt is suggestedd to adopt 0,25 value of A ULT coefficiient. One can n then obtainn in ULS:
Ppl , MCL 0,25 ex tw f y -charracteristic ressistance of coonnector for ULS steel crriteria
In SLS sstress criteriaa need to be fulfilled. It has to be guarantee g thaat in any poiint of shear connector (steel dow wel) yield sttrength is no ot excided. S tress pattern n calculated in i zone of shhear connecttor should take intoo account loocal pressurre of concreete on steel surface, heence reducedd stresses should s be determineed. General approach to maximum sstress calculaation was wo orked out in [Precobeam, Lorenc]. Proposedd formula is:
Ek ,SLS Ael
N M z G A I y tw I y
V Sy
Above, tthe first term m is shear sttress multipllied by Ael local l concen ntration factoor, the second one is normal sttress multipllied by βG global concenttration factorr. Unknown factors depeends only on n shape of steel connnectors and can be deteermined by m means of ex xperimental tests combinned with FE E analysis. Above m mentioned facctors are assu umed now too be of 7,00 and a 1,50 resp pectively. In FLS oone must prove that chaanges in prinncipal stressses (not redu uced ones lik ike in SLS) are small enough nnot to cause the t initiation n of fatigue ccrack. Speciaal care should d be paid to FLS when connectors c are placeed in the tennsile zone off cross sectioon, where in nitiated crack k can propaggate and leaad to total failure off beam. In caase of ELEM M bridges thiis is not the case. Stressees may be caalculated acccording to formula in SLS criteeria, with th he same conncentration factors f as th hey were callculated for FLS and he edge. Detaail category should be adopted ffor SLS apprroximating reeduced stres s by principaal stress at th safely asssumed as forr a gas cut ed dge, so 125 M MPa.
4.1.3 C Concrete failure crriteria Concrete failure criteeria governiing resistancce of compo osite dowels are presente ted in (P804 4)and it is recommeended to usee formulas given g thereinn. As mentio oned before shearing andd pry-out off concrete should bee verify in ULS. U Shearing resistance of concrete do owel can be estimated as: e 2 Psh 3 x ex 180
D
f ck 1 D
Es Asq ,1 Ecmm AD
AD 0,220 ex
2
49
where: ex spacing of doowels along web w Asq,1 reeinforcemennt in concretee dowel (sheaaring area) AD ddowel area Pry-out ccone criteria:
Ppo
35 f ck 1,5 1 D,i hpo 0,4 0,001 f ckk
hpo minnco 0,07 ex ; cu 0,13 ex
D,i
Es Asq Eccm AD ,i
AD , i hc ex
If ex / hpo < 5 then addditional redu uction factor iis calculated d: ex ex iff 4 0,2225 h h po po x ex 0,5 0,1 e x 1 iff 4 h po h po If ey / hpo < 10 (multi dowel stripss) then additiional reductio on factor is calculated: c
1 n 1 ey 1 1 n 10 hpo
y
Factor χy is calculateed very conservative andd for most off boundary conditions, c liike in POST Ts realized for ELEM M project, itt can be pro oved that theere is no reduction corrresponding w with adjacen nt dowel’s stripes iff its spacing is i at least 150 mm with w web thicknesss not greater than 20 mm m in concrette C50/60. ggested to bee omitted. In test evvaluation thiss factor is sug Indications like on draw. beloow (P804).
Figure 41: Indications for concrrete failure fformulas (P P804)
4.1.4 T Tests eva aluation an nd summ mary All POST Ts results were w evaluateed accordingg to EC4 pro ocedure. On this basis deesign resistaance of all specimenns were deteermined. Thee structure oof POST specimens wass assumed hhaving in baackground implemenntation of coonnection into segmentaal bridges, heence approprriate concrette coverage and strips spacing. Below exam mple of calcu ulations accoording to above describeed formulas for POST series A is given, annd comparisoon of estimaated and reall resistances for all POS STs are show wn. Short co onclusions about ulttimate resistaances of com mposite dow wels are pressented. Both concrete annd steel stren ngth were
50
obtained in tests. As concrete was tested on cube specimens fck is determined by dividing mean stresses fck,cube,mean by 1,18. Tangent modulus Ecm is determined on basis of fcm=fck+8 with accordance with EC2. Pmax [kN] 3653,5 3523,9 3340,8
Acronym PO. A 1/3 A 2/3 A 3/3
Pmax / dowel [kN] 608,9 587,3 556,8
PRk [kN]
PRd [kN]
501,1
400,9
δu [mm] 36,9 37,8 23,9*
δuk [mm] 33,2
upl 1.top upl 2.bottom [mm] [mm] 9,6 9,6 9,3 8,0 9,0 4,6
* tes ti ng wa s s topped a fter ma x force were rea ched not to a l l ow s peci men's tota l des tructi on
Figure 42: – Test evaluation for POST A acc. to EC4 Concrete resistance to shearing and pry-out:
e 2 Psh 3 x e x 180
D
Es Asq ,1 Ecm AD
250 2 f ck 1 D 3 0, 25 61,52 1 0,092 0,862 MN 180
200 0,000226 0,092 39,4 0,0125
AD 0,20 ex 0,20 0,252 0,0125m2 2
Ppo
35 f ck 1, 5 1 D , i h po 0,4 0,001 f ck 0,608
35 61,52 1 0,031 0,09251,5 0,452 MN 0,4 0,001 61,52
h po minco 0,07 ex ; cu 0,13 ex
min0,075 0,07 0,25 ; 0,11 0,13 0,25 0,0925 m
D,i
Es Asq Ecm AD ,i
200 0,000452 0,031 39,4 0,075
AD,i hc ex 0,30 0,25 0,075 m 2 If ex / hpo = 2,703 < 5 then additional reduction factor is calculated: 0,25 0,225 2,703 0,608 x 0,225 0,0925 Characteristic resistance to concrete failure:
PRk , conc min Psh ; Ppo min0,862 ; 0,452 0,452MN
Characteristic resistance to steel failure:
PRk , steel 0,25 ex tw f y 0,25 0,25 0,015 421 0,395MN Characteristic resistance of composite dowel is the smaller value from resistance to concrete and steel failure, hence: PRk min PRk , conc ; PRk , steel min 0,452 ; 0,395 0,395 MN PRd
PRk
v
0,395 0,316 MN PRd ,TEST 0,401 MN 1,25
(27 % smaller design resistance than real one)
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Proceedinng with furthher calculatiions analogiccal to steps shown abov ve, one can oobtain collecctive table with resuults for all tested speciimens (attenntion should d be paid th hat reductionn factor forr concrete resistancee, if more thhan 1 dowel’’s stripe is uused, is omittted, as it waas proved byy tests that th here is no resistancee reduction if i spacing bettween stripess is 150 – 30 00 mm): POST n no. PRd,t [kN N] PRk,s [kN] PR Rd,s [kN] 316 A tw15 1 1x3d 401 395 B tw20 1 1x3d 457 403 504 496 C tw25 1 1x3d 548 620 D tw20 1 1x2d 503 403 504 403 E tw20 2x2 2d e300 540 504 F tw20 2x2 2d e150 540 403 504
PRd,t/PRd,s 1,27 1,13 1,10 1,25 1,34 1,34
PRk,c [kkN] PRd,c [kN] PRd,t/PRd,c P PRd d [kN] 452 362 1,11 3 316 462 1,24 3 369 369 448 1,53 3 359 359 447 1,41 3 358 358 451 1,50 3 361 361 454 1,49 3 363 363
PRd,t/PRd 1,27 1,24 1,53 1,41 1,50 1,49
o for POS ST A (tw15) steel failuree governs the resistance,, what was proved p by It can bee seen that only tests alsoo. For web thickness t off 20 and 25 mm concrette failure is reached at llower total force, but analyzingg chart forcee – slip for POST P B tw220 one can notice n that in n reality steeel and concreete failure modes w were reached at the same time. This iis why reducction factor of o 1/1,15 = 00,87 was asssumed for calculatinng steel resistance (desccribed earlierr). Concrete resistance calculated c acccording to described formulas underestimaates real resisstance of aboout 25 – 50 %, % what is accceptable levvel of safety. It may be also conccluded that omitting o the reduction fa factor for tw wo parallel do owel’s stripees is fully ju ustified as safety maargin for all POSTs P is sim milar.
4.2 A Advanced d design of shearr connecttion girde er-slab In the foollowing, the bridge concept of a composite girder g grillag ge, abbreviaated “VTR” (German: Verbundtträgerrost), is described to t introduce the general structure s and d to make thee innovative idea clear and obviious. The maain construcctional elemeents includin ng their structural load-bbearing behaaviour are explainedd and the essential consttruction stepps are described. There are a several bbridges acrosss Europe applying the VTR cooncept. As th he constructiion progress in Poland iss farthest addvanced this particular bridge is chosen to bee described in n detail withhin this reporrt. The basiic concept of o VTR consists of a hiighest possible degree of o prefabricaation of thee different constructtional elemennts and a mo odular-type aassembly on n the construcction site (seee Figure 43). Quality of the inddividual elem ments can bee improved bby production n at the plantt which resuulting in econ nomic and time-depeending advaantages due to t the abolitiion of formw work at the construction c site. Prefabrrication of all consttructional ellements inclluding steel--componentss, cross-gird ders, carriageeway slabs and also reinforcement cages leads l to a sig gnificantly aaccelerated co onstruction progress p andd thereby to significant s reductionn of overall costs. c
Figure 43: Concept of system with w essentiall structural componentss
52
Yellow Green Blue Red
steel girder level cross girder level carriaageway levell reinfoorcement
steel girdeer + shear stu uds cross gird der + first in-situ concretee carriagew way slabs + seecond in-situu concrete reinforcem ment of all co onstructional al elements
Figure 44: Concept of system sh hown along the main co onstruction-ssteps The consstruction timee of composiite bridges iss in general determined d by b the constrruction of thee concrete carriagew way slabs. A formwork traveller is uused for lon ng bridges an nd the in-situu slab is con ncreted in different cycles. Thee weekly cycle is only achieved wiith short forrmwork travvellers so that several travellerss are needed to achieve a fast construuction progress. The herein presenteed bridge con ncept uses prefabricated concrette cross girders and carriiageway elem ments to min nimize workks to be acco omplished on site, soo that withouut formwork travellers orr falsework structures s a short construcction time is obtained. Thus, thee main innovvation of the VTR conceppt is based upon u the sign nificantly acccelerated con nstruction progress which is acccomplished along a the folllowing steps (see Figure 44):
P Prefabricationn of the steel girders (iincl. shear studs s and sh heeting eleme ments), concrete crossggirders and carriageway elements e P Placing of steeel-girders on n supports annd welding of o joints (yelllow parts) P Placing and arrangement a of prefabricaated cross giirders (green parts) A Assembly off reinforcem ment and conncreting of intermediatee sections annd gaps of the cross ggirders with in-situ i concrete (green paarts) P Placing and arrangement a of prefabricaated carriageeway slabs (b blue parts) A Assembly off reinforcemeent and conccreting of seections in-between the caarriageway slabs s with inn-situ concreete (blue partts) A Assembly of sealing, cover layer, canntilevers and completion of o bridge
53
All main constructionnal elements are prefabriccated, assem mbled on the construction c site and con nnected by means oof cast in-ssitu concrette. As the innovative constructio on technologgy requiress specific characterristics of the single structural elementts, these are described d wiithin the folloowing elaborrations: Longituddinal girders The longgitudinal girdders (see Fiigure 45-Figgure 47) are prefabricateed in steel aat the steel plant and transporteed to the construction siite. They cann be formed d as I-beams,, as closed aair-tight weld ded boxes and in U--form as upw wardly open boxes. Theyy fulfil the taask of distrib buting loads ffrom the cro oss girders and the carriagewayy slabs in lo ongitudinal ddirection by means of a composite cross sectio on. Three different cross sectionns result from m individuall construction n stages. The pure steel ggirder is useed to carry d its belongiing concretee. The steel girder togetther with firrst in-situ dead loads of cross girders and c slabs and thhe belonging g concrete concrete (green partss) is effective for dead looads of the carriageway (blue parrts). Finally, the composite girder inn combination with comp plete additionnal in-situ concrete is effective for loads froom bridge eq quipment andd traffic loads.
Figure 45: Steel gird der as box crross section
Figure 46: Steel-gird der on the co onstruction--site
Figure F 47: Lifting L of steeel-girders
Cross girrders The cross girders (seee Figure 48-Figure 50) are designed as reinforcced concretee girders wheereas prey favourable for certain applicationss. For the stressed cross sectionns are also possible andd eventually 1 cm, a heeight of 25 cm m and are specific pproject in Pooland the girders have a characteristiic width of 100 placed w within a patteern of 4.00 m onto the loongitudinal girders. g Caviities of 30 x 90 cm are assembled a which fitt into the paattern of sheear studs of the main girders. By grrouting the ccavities, a sh hear rigid connectioon between cross- and longitudinall girders is assured. Bettween the ccross girderss concrete sheeting elements aree arranged an nd fixed. In terms of load distrribution, the cross-girderss fulfil the fu unction to transfer loads from the carrriageway slabs to tthe steel girdders. They acct in transverrse direction only and in combinationn with the lon ngitudinal
54
girders set up the girder grillagee. During thee constructio on stages, tw wo different cross sectio ons occur. Loads off the carriageeway slab an nd in-situ cooncrete are caarried by thee prefabricatted cross-gird ders only. After com mpletion of the grouting g and in-situu concrete th he cross gird ders and carrriageway slaabs act in combinattion for both traffic-loadss and loads fr from bridge equipment. e
Figure 48: Cross girrder with reiinforcementt pattern an nd starter rebars for gapps
Figure 49: Cross-girrder on the construction c n-site
Figure F 50: Cross-girder C r on the steel-girder
First in-situ concrete The spacce on top of the steel-girders betweenn the cross girders g is gro outed with inn-situ concreete (green parts) in order to connnect the con nstructional eelements witth each otherr and to obtaain a uniform m level for positioninng the carrriageway slabs. Beforee grouting these gaps with in-sittu concrete,, specific prefabricated reinforccement cages are assembbled. The caavities for sh hear studs wiithin the cro oss-girders are groutted with speccial low-shrinkage mortaar while the remaining r gaaps are grouuted with con nventional concrete. Carriagew way slabs The carriiageway slabbs (see Figu ure 51) are aalso designed d as reinforcced concretee structures fabricated either at prefabrication plants orr at specific in-situ plan nts. Dead loads of the ccarriageway slabs are nly by speciific beam stu ubs for suppo ort and adjus ustment. Afteer all gaps transferreed to the crooss girders on are concrreted (blue components), c , impacting aactions resullting from traaffic and briidge equipmeent on the carriagew way slab aree transferred in both direections to th he cross gird ders and thee longitudinaal girders. Thus, thee slabs span in i two directions increasiing the consttructional efffectiveness.
55
Figure 51: Carriageeway slab wiith starter reebars for th he connection n to the girdders Second inn-situ concreete The remaaining spacee between th he carriagew way elementss is concreteed by the seecond in-situ u concrete (blue parrts). Therebyy, a continu uous surface for the seaaling and thee following bridge com mpletion is obtained and the loaad bearing beehaviour of the final staage is assureed. From thaat point on, the girder grillage iis a composiite construction, effectivve for loads from bridge equipment aand traffic loads. The concrete is applied on o the consttruction site and has to be reinforceed by additio ional longitu udinal and transversal reinforcem ment which can c favourabbly be set up as prefabricated cages. FFor this seco ond in-situ uction concreete is used. concretinng a conventiional constru Reinforceement Most parrts of the neecessary statiic and structtural reinforccement is asssembled witithin the prefabricated elements so that onlyy small amou unts of addittional in-situ reinforcemeent are needeed on the con nstruction site. Thee reinforcem ment patternss have to bee adapted to o each other in longituudinal and trransversal direction and especiaally to the paattern of sheear studs. In order to meeet demands of high accu uracy, the f be assembleed at the prefabricatio p on plant ussing specificc gauges. reinforcement can favourably d for a furthher acceleratiion of the Prefabriccated reinforccement cagees (accordingg to Figure 52) are used constructtion progresss.
Figure 52: Prefabriccated reinforcement cagge for acceleerated consttruction proogress For a succcessful impllementation of o this new ttechnology, important i an nd critical stru ructural detaiils have to be solveed. Due to high gradees of prefabbrication, th he requiremeents of accu curate produ uction are consideraably high forr all structurral elements. This e.g. ap pplies for thee reinforcem ment pattern within w the cross-girdders that haas to fit to the shear sttud pattern of the steell-girders. Thhus certain tolerancet measurem ments have too be included d and considdered during the design-p process in ordder to make inevitable geometriccal discrepanncies acceptaable for the cconstruction--process. Therefore, a veery careful an nd precise design annd planning process is crucial c for thhe successfull application n of the techhnology. Misstakes and inaccuraccies may lead to systemaatic failures aamong a larg ge amount of o prefabricatted elementss resulting in major constructionn problems. As A geometriccal patterns remain r the same and connstruction steeps repeat
56
in short terms, the tecchnology can n favorably bbe applied raather to large bridges thann to short and d medium span connstructions. For F such larg ge bridges cconsiderable economic advantages a rresult by a significant s reductionn of construction time. mportant annd critical deetails are desscribed whicch emphasizee the high im Certain im mportance off accurate productioon and sufficcient tolerancces for the coonstruction teechnology. Adaptionn of reinforceement-pattern n to shear stuuds The cross-girders proovide specifiic openings for the integ gration of sh hear studs asssembled on the steelgirder. T Thus, the reiinforcement pattern of tthe cross-girrders and th he shear stuud pattern haave to be adjusted to each otheer very carefu fully and acccurately. Furtther on, the prefabricated p d reinforcem ment cages der to assure a fast and efficient con nstruction have to bbe adjusted to the shearr pattern as well in ord progress. Patterns thaat are not ad djusted to eaach other maay lead to geeometrical ccollisions of structural elements which is veery time con nsuming andd thus expen nsive. As alll prefabricatted elementss are used numerous times, succh mistakes may have ggreat negativ ve consequen nces. Thereffore, it is esssential to adjust thhe (geometriccal) patternss of the diffferent constrructional elem ments to eac ach other and d to limit tolerancees to sufficiennt dimension ns (see Figuree 53-Figure 54).
Figure 53: Shear stu uds fitting in nto openingss
Figure F 54: Opening O for shear studs
Adaptionn of height off carriageway y slabs The carrriageway slaab elements are provideed with beam m stubs equ uipped with adjusting screws for support oon the cross girders. Theese stubs connsist of two U-profiles which w are heiight-adjustab ble by the screws inn order to com mpensate sm mall inaccuraccies of the co onstruction (see ( Figure 555-Figure 56)). In order to assure defined beaarings and th hus be able too estimate bearing forcess accurately,, only two beeam stubs for each ccross girder side are asseembled.
Figure 55: Openingss for shear studs s grouteed
Figure F 56: Adaption A of carriageway y slab
Summariizing, the VT TR technolo ogy is based upon the prrinciple to obtain o highesst possible degrees d of prefabrication for a significantlly acceleratted construcction progress. The efffectiveness of o saving constructtion time is achieved a for large span bbridges rather than short and medium m bridge consstructions. Due the usage of prrefabricated elements, itt is absolutely necessary y to meet deemands of very v close b the prefabbrication in outstanding o manner. m tolerancees which can be assured by
57
5 EFFECTIVE WIDTH AND STATIC BEHAVIOR To study the effect of dry joints on the load bearing behaviour of composite bridges, two different test series were performed. The test series at LTU focuses on shear lag, composite action and static capacity of the shear keys and the test series at RWTH Aachen focuses on the shear lag, composite action and the ultimate bending capacity.
5.1 Theoretical background A plate girder consists of a plated concrete flange and at least one beam-shaped (uniaxial) girder. Due to the flexibility of the concrete flange loaded with shear in its plane, the longitudinal stresses decrease along the width of the concrete flange (shear lag effect). In order to allow for static calculations using beam theory, the shear lag effect can be considered using a so-called effective width. This effective width is calculated by integrating the longitudinal stress of the real stress state and hence comparing it to a block-shaped stress state with the same center of gravity. The stress distribution can be calculated by plain stress theory or by plate theory, see next subchapter. The effective width is among others dependent on the following properties:
width to length ratio b / L type of loading level of loading distribution of the stiffness between flange and girder
EC4-2 provides design formulae for the determination of the effective width taking into account the most important of these factors. However, these formulae were derived for continuous slabs and are therefore not valid for bridges featuring segmented slabs. It is the aim of this chapter to derive formulae for effective widths for this kind of composite bridge. Effective width according to plain stress theory Cross section of bridges normally have a very high and thus stiff steel girder and a comparably thin concrete slab. For this circumstances the assumption holds that the concrete slab act like a disk therefore its and bending stiffness can be neglected. Hence, the stress distribution in a homogenous concrete slab can be derived by solving the differential equation for disks (Hake2007),(Lentrup2011):
x, y
4 4 4 2 0 y 4 x 2y 2 x 4
( 14)
using the shape function: n
1
n
2
cosh n y B n n y sinh n y cos n x An Cn sinh n y D n n y cosh n y
( 15)
with: -
n n
l
n 1, 2, 3...
As boundary conditions, different derivatives of the differential equation representing stresses and strains can be used. One example for such a derivative is the longitudinal stress:
x
2 y 2
( 16)
59
After solving the differential equation, the effective width can be derived according to:
bm,S
yx dx
( 17)
x
x ( y 0)
where τyx is dependent on the type of loading, see Table 7 for examples. Table 7: Examples of shear functions for different types of loads load type.
harmonic Load
uniform load
single load
system x
x
x
L
L
L
shear distribution shear function
R o sin( x)
8 k 1 sin( n x ) R o 2 ( 1) 2 n
sin( n x ) 4 R o n
The formula can be evaluated for ratios of effective width to geometric width (bm,S/b) over width to length (b/L), see Figure 71. However, if experimental tests are performed on down scaled test specimen, the bending stiffness of the concrete slab cannot be neglected any longer and must be taken into account. The stress distribution according to plate theory can be derived analogous. The interpolation of the effective width according to both theories is dependent on the stiffness ratio between steel girder and concrete slab. The formula is described in (Brendel1960) and (Kuhlmann2005) and reads: bm,S hc I io I a I co S io z st I a
bm , P hc 12 n0
3
n0 bm,S hc n0
Aa z st Aa
2
bm hc Aa 3 n0 bm hc 2 z st Ia bm hc 12 n0 Aa n0
( 18)
The differential equations were evaluated for both theories to obtain the stress distribution for different width to length ratios. These stress distributions can then be compared with the measured stresses.
5.2 Large scale laboratory tests 5.2.1 General In order to study the influence of the segmentation of the slab, various large scale tests were performed. The tests focused on the effective width, the composite action and the static capacity of the shear keys and the overall cross section. The specimen and the test set-up were prepared in a way which made it possible to perform tests simulating both a load situation in the field area of a continuous bridge, and a load situation above an internal support. The test specimen were designed with a real deck element, from a single span bridge over the river Rokån in Sweden, as model. The height (0,220-0,290 m) and the length (1,8 m) of the tested elements are the same as the size of prefabricated elements used on real bridges. The reasons to use the full height on the test specimen are to get the real dimensions of the shear keys and the studs, and to avoid discussions about how the scaling factors affect the results. The length of the specimen is in a real case governed by the distance between the parapet posts, which means 1,8-2,0 m in Sweden. The reason why the parapet post distance is governing the length of prefabricated elements is due to the benefits you can gain in both the production and the design stage, by using the same position of the post on each element.
60
5.2.2 T Test set-u ups Two test series were performed p att LTU and R RWTH Aacheen.
Tests at R RWTH Aach hen: The testss focused on o the effecttive width uunder saggin ng moment, see Table 8. Parameters under investigaation were the t size of the gap beetween adjaccent elements, the typee of loading g and the segmentaation comparred to a contiinuous slab. The test specimens consisted c off one weldedd steel girdeer (S 355) with w shear coonnectors (Ø Ø=22 mm, h=100 m mm) and 5 prefabricated p concrete deeck elementts made of C 35/34. Thhe dimension n of each prefabricated elemennt was 1,8 x 3,2 x 0,22 m m. The overaall length off a test specim men was 9 m and the span wass 8.8 m. In order to alllow for an uundisturbed developmen nt of the streess distributiion in the concrete slab, the wiidth of each concrete eleement was 3.2 m. The average a gap size was 5 mm m in the pecimens. Thhe geometry y of the steel girder and thhe concrete channel c is first and 2 mm in the other test sp depicted in Figure 577.
Table 8: Overview of o the perforrmed tests No. loadinng type
size of gap
span
Measurees
[mm]
[mm]
[m]
≤5
8800
5: 1.8 x 33.2 x 0.22
≤2
8800
5: 1.8 x 33.2 x 0.22
≤2
8800
5: 1.8 x 33.2 x 0.22
-
8800
1: 9 x 3.22 x 0.22
VT1
VT2
VT3
VT4
As injecttion mortar,, a non-shrinking mortaar fabricated d by the Geerman comppany Pagel was used (V1/50). This mortaar has a maaximum graiin size of 5 mm and compression sstrength of more m than 40 N/mm m² after 24 h, thereby allo owing for quuick erection times. The channel wass concreted in i 30 min. by 2 empployees of Pagel and req quired no com mpaction. The T tests werre performedd after 7 day ys and the measuredd compressioon strength of the morttar (fc=62 N//mm²) corressponded welll to the com mpression strength oof the concreete elements (fc=58 N/mm m²).
Figure 57: Geometryy of the test specimen, lleft: steel girrder, right: concrete chaannel
61
In order tto determinee the stress distribution d inn the cross section, s strain n gauges weere applied at a the steel girder annd on the top surface of th he concrete slab. In add dition, the slip between thhe steel girdeer and the concrete elements waas measured at five differrent cross secctions along the girder, seee Figure 58.
2
1
1 10
9
1 11
8
12 13 14
16 17
19 20
15
18
21
7
12
17
21
26
6
11
16
20
25
5
10
15
19
24
9
450
450
C-C
14
18
B BC-BC 450
320
13
CD-CD 900
I IV
V
SZ1
22
8
4
D-D 900
III
S5 6
SZ2
3
II
135 225 140
5
7
450
B-B
23
AB-AB 450
figure legend measurement of slip p
320
2
S4 3
I
LVDT 1600
4
A-A
AB-AB
35
S3
1
B-B
320
N
B BC-BC
strain gauge ation measurement of rota
320 320
S2
S1
C-C
CD-CD
D-D
measurement of displacement
A-A
45 50
[mm]
4500
Figure 58: Instrumeentation of the tests The load was appliedd as a point lo oad, 350x3500 mm, acting g in the centrreline of the bbridge.
Tests at L LTU: The test specimen consisted of two weldedd steel girderrs (S 355) with w shear coonnectors (Ø Ø=2x mm, h=100 m mm), three crross-beams and a four preefabricated concrete c deck k elements m made of C 35/40, 3 see h prefabricatted element was 1,8 x 3,,5 x 0,29 m. The overall length of Figure 599. The dimennsion of each the test sspecimen is 7.4 m with a span of 7 .2 m. In reall bridges, th he distance bbetween the girders is between 4 and 7 m deepending on the size of thhe bridge. Du ue to size lim mitations of tthe strongfield and the testing riig, the speciimen has beeen scaled doown in the transversal direction. d Thherefore, thee distance between the steel girrders is 3,0 m and onlyy the interiorr part of decck slab is coonsidered without the o the deck.. The effectts of the caantilevering parts are sstudied by using u FEcantilevering parts of calculatioons that havee been calibrated against the test results.
Figure 59: Drawingss of the test specimen [m mm]. Since sevveral differennt tests weree performed on one specimen, all loaad situations had to be kept below the failurre load resullting in non-destructive ttesting. Only y the last test situation, w when a shearr key was tested, waas performedd until a final failure.
62
During thhe tests strains were meaasured on thee steel girderrs, on top off the concretee elements and a on the reinforcement inside the elementss. The deflecctions were also a measureed on severaal positions, as well as the joint oopening betw ween the elem ments. Test set-uup no: 1 Figure 600 gives an illustration of test set-upp no 1. Two things are of o special innterest underr this load situation.. Firstly, willl the bridgee girders behhave as a co omposite section regardding to defleection and stresses? And how much m of the forces f will ennter the very y short concrrete slabs? S econdly, how w will the stresses inn the concrete be distribu uted regardinng to shear-laag?
Figure 60: Schematiic illustratio on of test sett-up no: 1 Test set-uup no: 2 Figure 611 gives a schhematic illusttration of thee load situatio on in this sett-up. One thiing that was examined was whether or not a bridge of this kind beehaves as an n ordinary co omposite connstruction in n the field sections. It was also checked if the t behaviouur would chaange after a couple of laarge load cy ycles. This he irregularitties in the drry joints willl be smootheened by local concrete was donee in order to find out if th crushing,, resulting inn a better fit after a a couplle of large lo oad cycles. Another A thingg that was stu udied was the shearr-lag effect, in order to find f out if a bridge of th his type behaves as a coonventional composite c bridge, reegarding to shear s lag in th he field areaa.
Figure 61: Schematiic illustratio on of test sett-up no: 2 The load was appliedd in two diffeerent ways. F First as a poiint load, 350 0x350 mm, aacting in the centreline of the briidge, and theen as two poiint load actinng straight ab bove the steeel girders. Thhe latter wass achieved by using a load distribbution beam.
5.2.3 R Results Tests at R RWTH Aach hen: Figure 622 and Figuree 63 depict the load dissplacement curves c of thee tests. Thee graphs reprresent the tested loaad and the exxpected ultim mate load resspectively, calculated with the real m material prop perties and neglectinng any safetyy factors. As can be seeen in Figure 62, the size of the gap ddoes not inflluence the initial stiffness but thhe ultimate bearing b capaccity, as stresss concentrattions in the ccontact areass between adjacent elements inccrease with higher h gap ssizes and aree governing initial failuree. Nevertheeless, both specimenns met the reqquired ultimate bearing ccapacity calcculated accorrding to EC44 under the asssumption of a contiinuous slab.
63
centre
Figure 62: Load displacement curve of tests VT1 and VT2 The increase of the concrete stress in the contact areas is also responsible for the lower bearing capacity of specimen VT 3 in comparison to specimen VT 4, see Figure 63. Furthermore, the segmentation leads to a more ductile behaviour as the concrete coverage failure in the contact area leads to a redistribution of stresses, whereas specimen 4 failed abruptly due to failure of the compression zone. The most significant difference between the segmented and continuous slab is the reduction of stiffness, which will be addressed in the following subchapter.
centre
centre
Figure 63: Load displacement curve of tests VT3 and VT4 During the tests, it could be observed that two different states have to be distinguished for composite girders with segmented slabs:
State A with open gaps State B with closed gaps
Furthermore, three different effective widths should be taken into account:
the effective width for displacement calculations, bm,V the effective width for stress calculations, bm,σ the effective width for ultimate bearing capacity, bm,T
A detailed explanation is given in the following chapters. Effective width for the calculation of deformations In order to evaluate the load-displacement curves derived from the experimental campaign, calculations were performed by integrating the moment-curvature relationship of the girders using different effective
64
widths, ssee Figure 64a. By usin ng an effectiive width bm,V of the sizze as the cooncrete chan nnel, very m accurate results could be obtaineed to describbe the initiaal stiffness, Figure F 64b. It can be taaken from Figure 644 that the use u of an efffective widtth for a hom mogenous sllab (bm,v = 24400 mm) results in a stiffness tthat is far tooo high and does not correespond to thee test results..
Figure 664: Compariison of calcu ulated displaacement currves with th he tests VT 1 and VT 2: a) entire curvee b) detail Table 9 summarizes values for bm,V, calculaated as secan nt moduli fo or load levells of 400 kN N. As this approachh is on the safe s side in all cases, aan effective width of thee size of thee concrete channel c is recommeended for displacement caalculations.
Table 9: Calculated stiffness of tests VT 1-V VT 4 VT 1 4
VT 2 4
Ii0 [mm ]
157719·10
1615211·10
bm,V [mm]
360
380
VT T3 4
VT 4 4
EC 4-2 4
166510·10
218592·10
267964·104
420
1060
2400
Effective width for thee calculation n of stresses, state A It could bbe observed during the teests that craccks developed from the edge e of the ellements at an n angle of 45 ° duriing the closuure of the gaap, see Figurre 65a. For this reason, the assumpption was meet that the elements act as a singgle slab for th he state in w which the gap ps are not clo osed. Figure 65b reveals the width to lengthh ratio, denooted as bi/L ratio, of thee test specim men (left han nd side) andd one singlee concrete element ((right hand side). 1,6 m
b i /L=0,18 8 m 8,8
1,6
m
b i /L=0,89 1,8
m
Figure 65: a) crack pattern p of sp pecimen VT T 1; b) b/L ratio the testt specimen aand a single slab The correesponding raatios of effecttive width too width of the slab are maarked in Figuure 66, wherre the bm/b ratios acccording to disk and platee theory are displayed. By interpolaating both thheories with formula ( 18) the bm/b curve forr the test specimen’s geo metry can bee obtained (ccontinuous linne, Figure 66 6).
65
Figure 66: Load disp placement curve c of testss VT3 and VT4 V This com mparison dem monstrates th hat for the geeometry of the test speciimen the infl fluence of the bending stiffness of the slab iss comparativ vely high for the separate element in case c that the gaps are nott closed. Figure 711 gives an exxample for the evaluatioon of the long gitudinal streesses in the concrete slab b. As the formula for the stress distribution is indep endent of th he current load, the meeasured stressses were normalizeed to allow for a comparison betweeen theoreticcal and meaasured valuess. By comp paring the measuredd values withh the calculaated shape o f the stress distribution d according too disk theory y, one can see that thhe stress disttribution derrived for a sinngle elementt is very closse to the meaasured data, see s Figure 71a. Furrthermore, thhe measured data lies bettween the sttress distribu utions calculaated accordin ng to disk and platee theory forr a b/L ratio o of the sinngle slab, Figure F 71b. This evaluuation streng gthens the assumptioon that the elements act as a single plattes in state A. A
Figure 667: Measureed stress sha ape for statee A comparred with a) different b//L ratios an nd b) disk and plate p theory for a b/L raatio of a sing gle element The effecctive width bm,σ is deriveed by an evaaluation of th he measured strains of thhe cross secttion under the assum mption of plaane deformattions and neggligible slip between girrder and slabb . These asssumptions could be verified for low load lev vels. Thereffore, the elasstic neutral axis of the coomposite cro oss section w was obttained by evaaluating form mula ( 19): was meassured and thee effective width zio z st
Aa z st Aio
z n Aa bm, st Aa Aa o b h zio hc Aa m, c no
( 19)
The resullts of the evaaluation are listed in Tabble 10, wherre bVT denotees the effectiive width intterpolated accordingg to ( 18) witth a b/L ratio o for a singlee slab.
66
Table 100: Calculated d stiffness off tests VT 1--VT 4 Test
VT 1
VT 2
VT 3
Ø
bm,σ [mm]
800
890
930
870
bi/L=0.89 Theory
EC4
Platee Theory
Disk Theory
bVT
bm,σ [mm]
650
18300
710
920
All meassured values were higher than the efffective width h according to EC 4-2 forr a single eleement and lie betweeen the calcuulated values according tto disk and plate p theory. For design purposes an n effective width acccording to EC C 4-2 is reco ommended foor a safe desiign. Effective width for thee calculation n of stresses, state B / ultimate bearin ng capacity At increaasing load levvels, the gap p closes and tthe stress disstribution in the slab apprroaches the shape s of a continuouus slab (Figuure 68a). Th his assumptioon is supporrted by the fact f that all sspecimens reeached the ultimate limit load acccording to EC4-2 E calcuulated under the assumpttion of a conntinuous slab b. Due to the advannced crack propagation, p the bearingg behavior of the slab reesembles a ddisk with a negligible n bending stiffness (Fiigure 68b). Based on tthe results of o the tests in terms off bearing cap pacity, no y EC4-2 coulld be observeed. significannt reduction of the effectiive width as proposed by
Figure 668: Measureed stress sha ape for statee B comparred with a) different b//L ratios an nd b) disk and plate p theory for a b/L raatio of a continuous slab b Tests at L LTU: Deflectioons The defleections can be b used to get an insightt in how stifff the entire system behaaves, and ind direct as a measurem ment of the composite c action. The defleections from m both test set-up s 1 and 2 inddicates a rathher linear beh haviour, but a sm mall part off the deform mations remains ( 0,5 mm, it is strrongly recom mmended to perform a Howeverr, if it is likelly that the jo non-lineaar FE-analyssis simulatin ng the gaps in the jointts that are closing c undeer an increassing load. Special ccare has to be b paid on th he edges of tthe concrete elements, which w can cruush in the co ontact line during cllosing the gaap. In Swed den, toleranc es allowing a maximum m gap of 0,44 mm have been b used successfuully using maatch cast con ncrete elemennts. In case oof negative bending mom ments, the moodel describeed in Euroco ode is not sui uitable for deefining the effective width of cooncrete flanges. The disstance betweeen the poin nts of zero bending mo oment, Le, cannot bee approximaated in the saame way. Le can never be b longer thaan the maxim mum distancee between the shearr studs, since the concretee element itsself cannot trransfer any lo ongitudinal ttensional forcces over a a FE-analyssis indicated that the strructural stifffness over an n internal joint. Labboratory tests as well as support (negative bennding momen nt) is rather cclose to the behaviour b off the steel secction itself. Therefore, T it is oftenn a quite goood approxim mation using only the stiiffness of thee steel crosss-section in the t global analysis. It there are doubts, wheether this appproximation can be used d on a speciffic bridge orr not, it is recommeended to perfform a sensittivity analys is studying the t effect on the momentt distribution n, where a part of thhe concrete iss included. EN 19944-2 5.4.2.3 (3) 3) The simpplified method that is described in this paragraaph should be b good enouugh also forr a global analysis oof multi spann composite bridges withh dry deck joints. But th he stiffness inn the supporrt regions, 15% of thhe span lenggth at each sid de of an inneer support, must m be chang ged. Eurocodde provides a stiffness
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for compposite sectionns with crack ked concretee, that is baseed on the mo oment of ineertia for the equivalent e effective steel cross-ssection (I2), including reeinforcement bars but exccluding conccrete in tensiion. Since we have nno longitudinnal reinforceement that crrosses the joiint, only the effective steeel girder cross-section should bee used to moodel the supeerstructure inn the global analysis. It is i also recom mmended to perform a small sennsitivity anallysis of the assumption a tthat 15% of the t span length from thee inner suppo ort should be treatedd as cracked..
3.1.2
Resistan nce of Cro oss-sectio ons
When thee resistance of o the superstructure is sttudied, the fo ollowing app proach has beeen used. If the conncrete deck is in tension, both bendinng moments and a normal forces f taken iinto considerration, the steel crosss-section is designed to take t the whoole load. Thiss approach will w of coursee add some steel to the support sections off a composiite bridge, approximateely replacin ng the area of the lon ngitudinal reinforcement, but wiith smaller distance to thee neutral ben nding axis. mpression, both bendin ng momentss and norm mal forces taaken into If the concrete decck is in com consideraation, then thhe resistancee is checkedd according to the rules given in EN N 1994-2. Iff only the elastic caapacity of thhe field sectiions is used in the desig gn calculatio on, then there re will be a rather big safety maargin for a faailure. For the llatter case, itt is assumed d that the cooncrete elemeents in comp pression behhave as an in n-situ cast deck. In tthe ULS thiss should be ok, o but regardding SLS and d especially FLS one shoould have in mind that the steel stresses migght increase a bit in the j oint sectionss, since a parrt of the loadd is taken by y the steel section onnly, before a possible joiint gap is cloosed. In mostt cases the neeutral bendinng axis will be b close to the upperr flange in a field section n. Therefore, the bottom flange f will often o be the ccritical detaill, together with the ddetails attachhed to the bo ottom flange that has to be checked fo or fatigue.
3.1.3
Concrete e elementt design
The conccrete design is performed d according to EN 1994 4-2 and EN 1992. There are some differences d comparedd to the desiggn of an in-situ cast deckk slab. Some of these are listed below w. The sheaar keys are of course a crritical detail in the design n of the elem ment. In the R RFCS-project ELEM, one type of shear keyys has been tested and eevaluated, wiith varying reinforcemen r nt layout. Th his type of shear keyy, Figure 355, has been proven p to bee suitable forr bridges wiith a girder sspacing ≤ 5,,0 m. The shear keyys are desiggned as a serries of overl rlapping male-female con nnections, aalways with one large shear keyy that distribuutes the load d in one direcction, and tw wo smaller shear keys in th the other direection, see Figure 122.
Figure 35: Type of Shear S key teested in labooratory and in single spa an bridges.
b deckss tends to vaary a lot duee to varying road profiless and due to o different Since thee design of bridge design traaditions in different d countries, the deesign of the shear keys will w also varry. The dimeensions of the elemeents will varry in three dimensions d (hheight, lengtth and width h). The distan ance between n the steel girders w will also varyy, as well as the t reinforceement layoutt in the elemeent. This secction shall bee seen as a
119
summaryy of recomm mendations an nd advices oof how shearr keys can be designed, rather than rules. r If a new typee of shear keey is inventeed or the reinnforcement layout l is chaanged a lot, it is recomm mended to perform ssome tests. The forcees transferred through th he shear keyss from one element e to an nother must bbe checked. Since the design off the deck slaab will vary from one brridge to anotther, it is stro ongly suggessted that a siimple FEanalysis is made for f each brridge, givingg the inforrmation needed to dessign the shear keys, transverse/longitudinal reinforcem ment etc. In an early dessign stage a simple s modeel, like the one o shown in Figuree 33, is oftenn accurate en nough to inveestigate the force f distribu ution betweeen the elemen nts. It can also be uused for the design of th he slab reinfforcement. In n Design Ex xample 1, thhis model is described more in ddetail.
Figure 36: FE-modeel to get the forces f in thee shear keyss (vertical deeformationss) m model, moddelling the fo orce transfer through the jjoints, is illu ustrated in One theooretical force equilibrium Figure 377 (suggested by Bo Westerberg, KTH H).
Figure 37: Illustration of force equilibrium e m model and the notation ns l gives Tests havve shown thaat an approacch assuming that only the inclined reebars carries the wholes load a results on the safee side, even though the most probab ble failure mode m in the tests a failu ure in the concrete cover. The capacity c to trransfer shear forces throu ugh the shearr key, is sugggested to be calculated c accordingg to the form mulas for incllined shear reeinforcementt in EN 1992 2-1-1 (6.13),
VRd ,s Asw f ywd sin Where
(1)
Asw = thhe area of thee shear reinfoorcement fywd = thhe yield stren ngth of the shhear reinforccement α = thhe inclination n of the shear ar reinforcem ment
mended that the t shear reinnforcement bars b in the sh hear keys maale-female connection It is stronngly recomm are overlaapping, see Figure F 38. This givees a more roobust constru uction in the ultimate lim mit state, sincce the shear keys will haave a post failure caapacity to traansfer forces even if the cconcrete coveer has been separated s froom the rebarss.
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apping shearr reinforcem ment. Figure 38: Illustration of overla
3.1.4
Steel dettail design n
The steell is designedd according to t EN 1994--2 and EN 1993. Essentiially the steeel is designed d as in an ordinary composite bridge. b The shear studs sppacing can however h givee a lot of probblems if it iss not done in a proper way. he transversee rebars in th he bottom The distaance betweenn the shear studs is goveerned by the spacing of th of the preefabricated element. e In the briidges construucted so far, the stud spaccing has been n 150 mm an nd the shear kkey depth haas been 60 mm. In thhe assemblinng stage, the new elemennt must be laaterally displaaced ≥ 60 m mm in order to pass the shear keyys on the forrmer elemen nt, see Figuree 39. If the shear s studs spacing s is 1550 mm, as well w as the spacing oof the transvversal rebars in the bottoom of the deeck element (Ø12 mm), tthe tolerancees will be about ± 222 mm (the rebar ribs taaken into acccount). If po ossible it is strongly s sugggested to inccrease the shear studds spacing annd the spacin ng of the trannsverse rebarrs, in order to o increase thhe tolerances.
Figure 39: Illustration of the lim mited toleran nces at the assembly a sta age. m in compputer program ms. It is quitee easy to makke a “preasseembly” of Today, allmost all draawings are made the bridge in the com mputer prograam, making ssure that therre will be no collisions beetween the sh hear studs and the reebars. Such a preassemblly is stronglyy recommend ded.
3.2 M Manufactu uring Compareed to a conveentional com mposite bridgee with an in--situ cast decck, the toleraances are verry limited. This makkes detailed control prog grams neces sary during the manufaccturing of thhe concrete deck. d The
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demand of higher precision on the location of the shear studs will limit the tolerances also for the steel workshop.
3.2.1
Deck elements
If a prefabricated bridge deck with dry joints is designed, it is strongly recommended to match-cast the elements in order to get a sufficient precision. The first element can be cast in an ordinary formwork. But from the second element and further, the previous element will act as formwork on one side of the next element. Some kind of surface treatment (oil, plastic film etc.) is recommended on the element that is acting as formwork. All elements shall be numbered, in order to assure that they are mounted on the bridge in same order as they are produced. It is also recommended to perform additional controls of the dimensions of the formwork both before and after the wet concrete enters the formwork, to assure that the formwork is rigid. Some of the additional controls that are recommended are listed below. Before casting -
Levelling of formwork
-
Geometry of the formwork (height, width, length + both diagonals checked)
-
Concrete cover thickness
-
Accuracy of reinforcement location in critical sections. (rebars in shear keys, rebars crossing the steel girder and the row of studs)
After casting -
Levelling of the element (assures that deformations do no occur during casting)
-
Geometry of the element (height, width, length + both diagonals)
-
Position of the rebars in the bottom of the slab, crossing the steel girder and the row of shear studs.
-
Joint gap between the new element and the previously cast element.
-
Smoothness of joint surface. If there are irregularities that are not negligible, these should be grinded away.
It is important that the mean value of the element lengths results in a value very close to the desired distance. The number of elements and the tolerances between rebars and shear studs will influence the maximum allowed tolerances for the elements as a group. As an example, for a bridge with a desired element length of 1,800 m, where the real length of all elements are 1,797 m. All elements might fulfil the tolerances for single elements, but as a group of 10 elements the total length will decrease with 30 mm. If the studs spacing 150 mm is used, the last elements will be very hard to get in their right position, even with a perfect positioning of the shear studs.
3.2.2
Steel
The steel tolerances are also more limited in this type of bridge, compared to a composite bridge with an in-situ cast deck slab. All details in the intersection between the concrete and the steel must be in right place, to avoid problems in the assembly stage. The position of the shear studs is crucial. To minimize the risk of collisions between rebars and shear studs. The relative positions of the shear studs should be measured within one group (element), and also
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between the groups. It is very important to avoid that the same error is repeated when the studs are positioned, such an error will soon be larger than the tolerated values.
3.2.3 Waterproofing For the integrity of the water proofing we must distinguish between elements produced using match casting or traditional formwork. We must also distinguish between regions above and near intermediate supports versus spans along which the deck is in compression. Traditional water proofing is made of at least one layer of 5 mm thick bitumen impregnated polyester/glass-fibre felt welded to the deck. This membrane is the actual water proofing on top of which different kinds of asphalt products are applied. It is also clear that it is the deck-to-deck opening induced by traffic at low temperature conditions that is most detrimental. Slow displacements caused by for instance temperature, concrete shrinkage, support settlements, etc should have a negligible influence. Match casted elements are preferred because the traffic induced joint opening will in general be smaller at all positions along the bridge compared to elements produced by traditional techniques. Indeed if match casted elements are used on a single span bridge the membrane and the asphalt can be applied to the deck surface without any extra arrangements over each deck-to-deck joint. This requires that the traffic induced opening displacement does not vary more than 0,2 mm under low temperature conditions. If on the other hand the elements have not been match casted displacements greater than 0,2 mm are quite likely, even for a single span bridge. In this case the elongation of the membrane must be distributed over a longer length in a similar manner as described for intermediate supports below, but the elongation length can be kept smaller. The trickiest part with regard to waterproofing is by far the regions around intermediate supports of multi-span bridges. Here, some recommendation on how to use traditional water proofing over deck joints in such regions are given (confer Figure 40):
If just one single waterproofing membrane is used, some kind of artificial de-bonding between the membrane and concrete as well as between the membrane and asphalt must be present to substantially improve the fatigue resistance. The total width of such regions must be about 20 cm. For this solution to work it is probably also necessary to artificially control the cracking of the asphalt layer such that it is more or less located over the deck joint opening rather than over the edge of the de-bonded region. Otherwise, the de-bonding strip cannot prevent the crack from growing into the membrane.
If two or more water proofing membranes are used the above statement holds but the de-bonded regions must increase in width when going from the top towards the concrete surface. Otherwise, the de-bonded region will not be able to stop a crack in an upper layer from growing into an underlying one. This reasoning holds as long as the cracks are formed and propagates from the top and downwards, which is the most likely scenario over an intermediate support of a bridge deck. The width increase should not be less than 5 cm, even if these regions can be placed with great accuracy.
One cannot rely on natural de-bonding between the material layers. De-bonding will not occur, not even between the concrete and the membrane, before the of asphalt layer cracks right through over the deck-to-deck joint. Best practice is to artificially ensure that the asphalt cracks in line with the deck-to-deck joint such that the already de-bonded region below can stop the crack from propagating into the membrane. This can, for instance, be achieved by putting some kind of rubber based product in the asphalt layer right above the joint and towards the top membrane. In combination with mastic asphalt, which in itself is water tight, this rubber can act as a fist seal preventing water an dust from penetrating into the de-bonded region below.
The actual de-bonding can be achieved by any practical means. But in order to promote rapid assembly and sufficient quality it should preferably be built into the membranes themselves. Here, product development in collaboration with some membrane manufacturer may be needed. The
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meaans used for de-bonding must in all cases ensuree that the materials doess not stick or o bond to eachh other as tim me passes.
Masstic asphalt iss preferred in n contrast to traditional asphalts, a as this t product iis in itself water w tight, whicch will effecttively localisse the water pproofing pro oblem to each h deck joint. This will alsso make it easieer to protect the de-bonded regions ass stated abov ve.
Figure 40:
ngement of water prooffing membrranes and de-bonded reegions (deno oted Db1, Arran Db2 and a Db3) ov ver a deck-too-deck joint at an interm mediate suppport.
For a joiint, producedd following the above rrecommendaations, the expected life time is mo ore than 2 million cycles of 2,0 mm m joint dissplacement aat -20°C. If th he temperatu ure is higher tthan -20°C, the traffic induced jjoint openinng is smallerr than 2 mm and/or the de-bonded regions r are llonger than 20 cm the number oof cycles to failure f will be much greatter. There aree two major drawbacks with the aboove solution.. The first iss that deliberrate de-bond ding is not allowed iin the presennt regulationss of some Euuropean coun ntries. It is su uspected thatt the de-bond ded region may grow w in size whhen for instaance passed bby heavy veehicles. The second is thhat water, du ust and all kinds of pollutions may m penetratee down to thhe membranee as soon as the asphalt cracks. Espeecially the water maay increase the de-bond ded region i f it repeated dly freezes to t ice. The first problem m can be counteraccted by usingg a thicker asphalt a layerr than usual, and the seco ond by usingg some kind of rubber sealing ass described above. a
3.3 A Assembly y 3.3.1
Steel
The sheaar studs andd their positions are bellieved to bee the most critical c detaiil on the steeel in the manufactturing stage. The alignm ment of the stteel girders is also very important. E Experiences from real bridges hhave shown that t there is no idea focuusing on the alignment, before b the steeel girders are in their final posiition (after laaunching/liftting). Laterallly adjustablee cross stays can be usedd to adjust th he position of the girrders. Attachhment points for such crosss stays shou uld be consid dered in the ddesign stage.
3.3.2
Concrete e elementts
The elem ments must be handled caarefully durinng the transp port and asseembly, in ordder to avoid damages. It is stronngly recomm mended to av void right-anngled cornerr wherever possible. p Thiis can for ex xample be done by uusing spliness (45°) in thee formwork. The elem ment is often handled by a crane, thatt lowers the element in th he displacedd position (> the depth of the shhear key), beefore the elem ment is slideed into the final fi position n, see Figuree 41. In order to make
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sure that the joint gaap has been closed as goood as possiible, it is reccommended that the elements are pushed toogether in onne way or an nother. Differrent techniqu ues have beeen tested. Sm mall portable jacks can be suppoorted by the shear s studs and a be used to push the elements tog gether. In shoort bridges, where the steel girdders often aree cast into th he back wallss at the abuttment, bolts can c be used to pull the back b walls against thhe deck elem ments, clampiing them tog ether, beforee the steel girrders are castt into the bacck walls.
Figure 41: Assemblyy of deck eleements
3.3.3
In-situ ca ast concrrete
It is hardd to perform a good com mpaction of thhe concrete in the in-situ u cast channeels, by using g concrete vibrators. Therefore, it is recomm mended to uuse Self-Com mpacting Co oncrete (SCC C) for the in n-situ cast channels.. The conccrete can be injected thro ough injectioon holes. In the tests as well as in thhe real bridg ges, Ø100 mm injecction holes have h been used, with a sppacing of 0,6 6 - 1,2 m. When W the channnels are injected it is importannt to make suure that the air a can escappe. Thereforee, in addition n to the injecction holes, air a release holes witth smaller diaameters are recommende r ed, Ø16 mm s300 mm have been usedd successfullly.
3.4 D Dry joints s vs. wet joints j Concrete bridge deckks with dry joints j are m most suitable for shorter bridges. If tthe technique shall be used on llonger bridgees, there willl be a need oof wet joints in order to make m it posssible to zero the errors that havee been summed up. If ordinarry wet joints are used, it will w not be ppossible to in nstall the water insulationn layer almosst directly. This meaans that the bridge b canno ot be taken innto traffic ass fast as in th he case of a bridge with dry joints only. c witth varying curvature, c deck width etcc., a solution n with wet If the geoometry of thhe bridge is complex, joints willl probably be b more suitaable, since it is possible to o adjust the elements e a biit in each joiint. If there iss a need to replace r a brid dge quickly, or to build a bridge in an n area with rrestricted acccessibility (above raail way trackks etc.), a brid dge deck witth dry joints can c be a goo od solution.
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3.5 Summary Prefabricated concrete bridge decks with dry joints are well suited for single span bridges. Due to gaps in the joints the bridge might behave a bit weaker than a similar bridge with an in-situ cast deck under lower loads, in SLS. This means that if the deflections are strictly limited or if the free-height below the bridge is limited, one should take into account that the deflection might be a bit higher. This should also be considered in the fatigue calculations of the steel girders. One way to deal with this is to decrease the area of the effective concrete width. It is possible to implement the same technique on multi span bridges, but there are some important factors that should be considered. Since the steel cross-section is designed to take the whole load, when the deck slab in in tension, this type of bridge will require slightly more steel in the support sections of a composite bridge. If an economic analysis is made, the material costs will probably be higher compared to a case with an in-situ cast deck slab. The benefit with this type of bridge is however not lower material cost, but rather fast assembly, better working conditions, absence of formwork, lower road user costs etc. It is also necessary to make sure that the pavement and the water insulation layer can withstand the opening of the joints over an internal support. Tests have been done at the Royal Institute of Technology in Sweden (KTH) in order to study the fatigue capacity of the insulation layer, see chapter 3.2.3.
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4 DE ESIGN EXA AMPLES 4.1 E Example 1 – Desig gn of she ear key re einforcem ment This exaample presennts how thee forces in the shear keys k can bee estimated,, and how the shear reinforcement can be designed. omposite briddge with a frree width of 7,0 m. The ssuperstructurre is made The bridgge in the exaample is a co of two stteel I-girderss with a spaccing of 4,0 m m. Prefabricaated deck slaabs are placeed on top off that. The deck slabbs are made of o C40/50 co oncrete. The prefaabricated eleements have a varying plaate thicknesss with 300 mm m in the cenntre of the brridge, and 216 mm in the thinneess part near the edge beaams, see Figu ure 42. The length l of a siingle elemen nt is 1,800 m.
ucture geom metry Figure 42: Superstru he deck slab are presenteed in Figure 43. 4 In one The dimeensions of thhe shear keyss in the interiior part of th longitudiinal directionn two small shear s keys trransfer the sh hear forces between b the elements, in n the other direction a single largger shear key ys transfers thhe forces.
ns Figure 43: Shear keyy dimension h two types of o rebars, seee Figure 44.. The SXThe sheaar keys in thiis bridge aree mainly reinnforced with rebars accts as shear reinforcemen r nt, and the E E-rebars ensu ures the distrribution of thhe forces between the SX-rebarrs. The E-rebbars can also take care off horizontal teensional forcces in the sheear keys.
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Figure 44: Shear keyy reinforcem ment. In order tto find the foorce transferrred by the shhear keys, a simple FE-m model of a seeries of deck k elements are madee. The elemeents are modeelled as simpply supported d by the steeel girders, witith no interacction at all in the joinnts between the elementss, despite in tthe points off the shear keeys. In this ppoint a rigid element e is used to trransfer the force f from on ne side of thhe joint to an nother. In thee model beloow, see Figu ure 45, the element iin the middlee is loaded. On O one side of this element all jointss are modelleed with one shear s key, on the othher side two shear keys are a used. Thee rigid links between the elements aree placed in the middle of each shhear key, andd illustrated in Figure 455 by the elem ment numberss (25-38).
ation of sheaar key forcess. Figure 45: FE-modeel for estima The characterristic load The midddle element is loaded wiith traffic loaad model 1 and 2 (LM1 & LM3). T values froom EN 19911-2 are used, which meanns that all -ffactors are seet to 1,0. Thee worst load situations for the laarge shear keey and the sm maller shearr keys are presented in Figure 46 andd Table 3. LM1 L is the worst trafffic load for both cases.
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Table 3: Load positiions for max ximum forcee in the two types of sheear keys.
Axle load ds Axle load d Lane loa ad 1 Lane loa ad n VSd
2x300 kN 400 kN 9,0 kPa 2,5 kPa = =1,5 x=
Large e shear key LM1 1 LM2 109,8 8 104,2 9,5 5 0,2 2 179,1 1 156,3
Small shear s keys LM1 LM2 73,6 75,4 5,5 0,1 118,8 113,0
m forrce in the tw wo types of shear s keys. Figure 46: Load possitions for maximum k is suggested to be caalculated acccording to The capaacity to transsfer shear forrces throughh the shear key, the formuulas for inclined shear reiinforcement in EN 1992--1-1 (6.13),
VRd , s Asw f ywd sin Where
(2)
Asw = thhe area of thee shear reinfoorcement fywd = thhe yield stren ngth of the shhear reinforccement α = thhe inclination n of the shear ar reinforcem ment
Large sheear key
Asw 8
Ø2 4
Smalll shear key
8
12 2 4
905 mm²
Asw 6
Ø2 4
6
12 2 4
679 mm²²
fywd = 5000/1,15 = 435 MPa
fywd = 500/1,15 = 435 4 MPa
α = 60°
0° α = 60
VRd ,s Asw f ywd sin 341 kN
VRd ,s Asw f ywd sin s 256 kkN
VSd 1779 kN
VSd 119 kN
V Rd , s V Sd
V Rd , s VSd
OK K
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OK
4.2 E Example 2 – Desig gn in sec ctions witth hoggin ng mome ent In this eexample a 265 2 m long g five span bridge is studied, s Forssjösjön Briddge. This brridge was constructted in 2011 as a an ordinarry compositee bridge, witth an in-situ cast concrete te deck. Cut outs from the generral drawing can c be seen in n Figure 47 aand Figure 48. 4
Figure 47: Forsjösjöön Bridge elevation.
Figure 48:
Typiccal cross-secction
F Figure 49: Picture P from m launching in 2011.
t differencces in the de sign at intern nal supports,, and to givee an indicatio on of how In order tto illustrate the the steel weight is afffected, a co omparative ddesign has beeen done witth a prefabriicated deck. The steel cross secttion is a hybrid girder, with w S 460/S 4420 in the flaanges and S 355 in the w web plates. The stiffn fness of the superstructur s re in the suppport section ns (15% out in the spanss) are modelled as the steel secttion only. Thhis implies th hat the supp ort moment will decreasse in comparrison to an in n-situ cast deck (55% in this caase), in which the reinfforcement arrea is included in the stiiffness in th he support sections. However, thhe sectional modulus wiill also decreease (25% for the uppeer flange), reesulting in han in a simillar in-situ caast bridge. steel stressses that are far higher th In the case with an inn-situ cast brridge deck, tthe deck is casted c in 9 stages, givingg step wise composite c dge. In case oof a prefabriccated deck, the t total loadd from the deeck slab is action in different parrt of the brid acting onn non-compoosite cross-seections. Sincce no composite action iss assumed inn the supportt sections, momentss and normal forces can just be summ med up from different loaads. In Tablee 4 the sectional forces in supporrt section 3 are a summarissed. Table 4: Cross-sectional forces at support 3 (with a preefabricated concrete deeck). Section x = 103,,169 m Steel Conc. dec ck Railing + walkway Pavement Shrinkage e Temp. grad Support s settl. LM1 Special ve ehicle Breaking load
Characteristic loa ads M [MNm] N [MN] 0 -2,43 -14,23 0 -0,40 0 -3,28 0 -2,74 0 -2,22 0 -0,72 0 -12,84 0 -14,44 0 -0,01 1,2
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ULS LF 1,2 1,2 1,2 1,32 1,2 0,9 1,1 0 1,5 0,68
Dim. Loads Mdim [MNm]] Ndim [M MN] -2,92 0 -17,08 0 0 -0,49 0 -4,33 0 -3,29 0 -2,00 0 -0,79 0 0 0 -21,66 0,81 -0,01 0,81 -52,55
Below, the steel cross-section in the in-situ cast bridge is compared to the cross-section in the prefabricated alternative. In the in-situ cast alternative we assume 1% reinforcement in the concrete deck slab. In order to get a similar utilization ratio in both alternatives, the thickness of the upper flange has in this section been increased from 50 mm to 64 mm, in the case with a prefabricated deck. Section x = 103,169 m web t [mm] web h [mm] b.flange t [mm] b.flange w [mm] t.flange t [mm] t.flange w [mm] web.red. A [mm2] I [mm4] CG [mm] conc. nL or n0 Aconc [m2] I [m4] CG [mm] eCG [mm] Area [mm²] Ix [mm4] Wtfl [m³] Ww.t [m³] Ww.b [m³] Wbfl [m³]
ELEMBridge 22 2386 50 1000 64 750
Section x = 103,169 m web t [mm] web h [mm] b.flange t [mm] b.flange w [mm] t.flange t [mm] t.flange w [mm] web.red. A [mm2] I [mm4] CG [mm] conc. nL or n0 Aconc [m2] I [m4] CG [mm] eCG [mm] Area [mm²] Ix [mm4] Wtfl [m³] Ww.t [m³] Ww.b [m³] Wbfl [m³]
-1543 -6,3E+05 1967
1264 0,1489 0,17035 -0,1348 -0,1420 0,1436 0,1378
tfl = 396 MPa bfl = 387 MPa
In-situ cast deck 22 2400 50 1000 50 750 -595 -3,6E+05 1990 100,0 1,565 0,0120 -195 1200 0,1554 0,18875 -0,1573 -0,1641 0,1510 0,1452
tfl = 395 MPa bfl = 379 MPa
Rotations The steel girders are precambered for the weight of the concrete and the steel. In theory, there will be no joint openings from these loads if the elements can be pushed together after installation. All loads that are applied after the injection of the channels will however contribute to the rotations at the internal supports. The theoretical joint openings at the internal supports are estimated by the equation below.
jo int
M
mean
/ Wtop. fl eCG hconc Lelement E eCG
(3)
Mmean
= mean value of the moment along the length of one element, in this case 50% of the length of the elements on each side of the internal support.
Wtop.fl
= elastic section modulus at the upper side of the upper flange.
eCG
= vertical position of the neutral bending axis, e = 0 at the top of the top flange
hconc
= concrete thickness, 0,290 m in this case.
Lelement
= element length, in this case 1,800 m
When the joint opening is calculated for ULS all moments are summarized, except the moments due to steel and concrete dead loads. In the fatigue limit state (FLS) only the moment caused by the fatigue vehicle is taken into account (FLM-3).
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Maximum m joint openiing – ULS
jo int
32 ,6 / 0 ,1348 12664 350 1,8 210 100 3
1264
= 0,0027 m
Maximum m joint openiing – FLS
jo int
4 ,7 / 0 ,11348 12664 350 1,8 210 100 3
1264 1
= 0,0004 m
The jointt openings are a most interesting for the sustainaability of thee water insuulation as well w as the pavemennt. Fatigue tessts performeed at KTH inndicates that the water inssulation layeer can resists 2 million cycles wiith a displacement ampliitude of 2,0 m C before failure. This meeans that theere will be mm at -20°C no probleem for the water w insulation layer to rresists the fatigue it will be exposed tto during itss technical lifetime oof 40 years. This is true, if the assem mbly recomm mendations giiven in sectioon 3.3 are fo ollowed or at least coonsidered veery carefully if any alternnative design is used.
Conclusioons The geneeral conclusiion is that it will be neecessary to add a some steeel in the uppper flangess near the internal ssupports, if a bridge is designed d withh prefabricatted deck elem ments with ddry joints instead of a concrete deck cast onn site. At thiss specific briidge, it is neecessary to ad dd some steeel to the upper flanges 25% out in the spans from the intternal supporrt. The total steel s weight of the solutioon with an in n-situ cast deck is 465 ton, annd the addiitional steell needed in the solutio on with a pprefabricated d deck is approxim mately 21 tonn. The assem mbly joints in i the steel girders are in this case optimized for an a in-situ casst bridge. It is i possible to lower the amount of additional steel in thhe alternativ ve with a preefabricated ddeck, if the joints are optimizedd for this typpe of construction insteadd. The jointt openings duue to the crosss-sectional rrotation will be no probleem in ULS oor FLS.
4.3 E Example 3 – Extra a control program m If a desiggn solution with w a prefab bricated conccrete deck, with w dry jointts, are chosenn instead of an in-situ cast deckk. It is necesssary to add an extra conntrol program m to make su ure that the limited tolerrances are achieved. An extra control proggram is preseented in this m has been example. This conttrol program developeed from a controll program successfuully used inn the constru uction of a Swedish single span bridge, b AC 1684 1 outside Norrfors (2002), see Figure 50. This T control program has been im mproved in lin ne with new experiencces, that havve been gatthered from other eleement bridgees, literaturee studies, as well as tthe experiennces of constructing the large scaale test speecimen in the t ELEMproject. Figure 50 0: AC 1684 after a assembbly.
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Bridge daata The bridgge which thiss example is based on is a single span n composite bridge, withh a span leng gth of 28,8 m. The ssuperstructurre consists of o two steell I-girders with w a spacin ng of 4,0 m m, and a prefabricated concrete deck on topp of that. Th he steel girdder ends are integrated into i the backkwalls by in n-situ cast joints. Thhe deck conssists of 16 elements + tw wo prefabricaated end elem ments, whichh consists of back- and wing-wallls. All transvversal joints are complettely dry, and d in-situ cast concrete is oonly used to create the connectioon between the t steel gird ders and the elements. Th he general geeometry of th the bridge is presented below in Figure 51.
Figure 51: Geometryy of Norrforrs Bridge. Extra con ntrol prograam The descrribed control program beelow containns only tasks that occurs specific s on coomposite briidges with prefabricated deck eleements with dry joints.
In concreete element workshop w -
leevelling of thhe formwork k fformwork geometry innitial joint gaap, each join nt measured bby test assem mbly sstorage environment (hum midity, tempeerature, etc.)
workshop In steel w -
pposition of eaach group off shear studs pposition of reeference mark ks innclination off the upper fllanges ssmoothness of o upper flanges (no weldd spatter)
On the brride site -
jooint gaps aftter assembly (from the uppper surface)) cconcrete injecction in the open o channells over the stteel girders ccross-measurrement of steel girders.
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Tolerancces and contrrol methods In this section somee of the most importantt controls arre described d more in deetail. Focus is on all interactioon Levellingg of formworkk. The prevvious cast eleement acts ass formwork oon one side of the new element. e Thee former is adjusted in order to gget a good connection c between the fformwork an nd the element. Figure 52 illustrates how the levelling is measureed. The toleerance of d is set to ± 2 mm. In n connectionn to the casting, the deformattions of the formwork f shall be checkked, in order to not exceeed the tolerannce. A levellling of the formworkk and the eleement shall be b performedd in the liness correspondiing to the steeel girder centre lines, both befoore and after casting.
n of levelling tolerancess Figure 52: Definition Formworrk geometry It is of hhighest impoortance that the t plan geoometry of the deck elem ments is correect. Especiallly in this case, wheen there is a plan radius of the bridgge centre line. Figure 53 shows the ddistances thaat shall be checked bbefore the ellement are caast. The toleerances of L1 1, L2, D1 and D2 are in tthis case all set to ± 3 mm.
Figure 53: Geometryy control distances Initial joiint gap The initiaal gap in eacch joint shall be measureed at the con ncrete worksshop, by doinng a test asssembly. In this stagee, without anny prestressin ng forces on the element, the mean value v of the ggaps are allowed to be 1,0 mm. The gap caan be measurred by usingg feeler gaug ges. If a feeeler gauge oof 0,30 mm cannot be
134
pushed innto the joint, the joint gap is consiidered as 0,0 mm. The joint gap iss measured at several positions along the jooint, see Figu ure 54, both ffrom the top and the botto om side.
Figure 54: Measurem ment positio ons for jointt gaps. Joint gapps after assem mbly The jointt gaps are alsso measured after the asssembly. In th his case the gap g is measuured only fro om above, since it ccan be quite hard h to get to the point uunder the briidge. At this stage after tthe element have h been pushed ttogether duriing the asseembly, the m mean value of the joint gap shall be ≤ 0,40 mm. The maximum m allowed gap g is ≤ 1,5 mm locallyy over a maaximum disttance of 1,00 m. The joiint gap is measuredd at the posittions as preseented in Figuure 54. Shear stuuds positionss The tolerrance of a sinngle shear stu uds is ± 5 m mm. But it is of o highest im mportance tha hat the same mistake m is not repeaated again and a again. Th herefore, thee position off the first sh hear studs inn each group p shall be checked, as well as thhe distance between the ffirst and the last l shear stu ud, see Figuree 55.
uds tolerances. Figure 55: Shear stu o steel girdeers Cross-meeasurement of In order to make surre that the steel s girders are in the right r position ns, in relatioon to each other, o and aligned ccorrectly, a cross-measur c rement shalll be done. The T measurement points shall be maarked and checked already in thhe steel worrkshop. On tthe site, meaasurements of o these disttances can be b used to adjust thee girders intoo the right position. p Figgure 56 show ws the theoreetical distancces, and a fo ormula for calculatinng the necesssary displaceement of girdder B. The fo ormula is only valid for thhis specific bridge. b
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s cross-m measuremen nts. Figure 56: Plan for steel Concretee injection The fillinng ability of the t injected concrete c shaall be establisshed by full scale tests.
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5 REFERENCES Culmo, M. (2009). Prefabricated Composite Bridges in the United States including Total Bridge Prefabrication. Workshop on Composite Bridges with Prefabricated Deck Elements. Stockholm, Sweden.
Berthellemy, J. (2001). “’Composite Construction - Innovative Solutions for Road Bridges”, Proc. 3rd International Meeting on Composite Bridges, Jan 2001, Madrid, Spain. Berthellemy, J. (2009). “French Experiences from Prefabricated Deck Elements”, Proc., Workshop on Composite Bridges with Prefabricated Deck Elements, March 4th 2009, Stockholm, Sweden Collin, P., and Johansson, B. (1999). “Wettbewerbsfähige Brücken in Verbundbauweise“, Stahlbau, vol 68 Heft 11: 908-918 (in German) FHWA, (2009). Connection Details for Prefabricated Bridge Elements and Systems, FHWA-IF-09-010, USA FHWA, (2010). “Field Cast UHPC Connections for Mopdular Bridge Deck Elements”, FHWA-HRT11-022, USA Harju, T. (2009). “CASE – Laisentianjoki Bridge”, Proc., Workshop on Composite Bridges with Prefabricated Deck Elements, March 4th 2009, Stockholm, Sweden Hällmark, R., Collin, P., and Stoltz, A. (2009). ”Innovative Prefabricated Composite Bridges”, Structural Engineering International, vol. 19, no 1: 69-78 Seidl, G. (2009): Composite Element Bridges, Workshop on Composite Bridges with Prefabricated Deck Elements, March 4th 2009, Stockholm, Sweden (ISBN: 978-97-7439) Seidl, G. Braun, A. (2009): VFT-WIB-Brücke bei Vigaun - Verbundsbrücke mit externer Bewehrung, Stahlbau, vol 78 Heft 2 (in German)
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European Commission EUR 25897 — Composite bridges with prefabricated decks (ELEM) Luxembourg: Publications Office of the European Union 2013 — 137 pp. — 21 × 29.7 cm ISBN 978-92-79-29156-2 doi:10.2777/79809
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KI-NA-25897-EN-N
Bridges are of vital importance to the European infrastructure and composite bridges already became a popular solution in many countries and a wellestablished alternative to concrete bridges. To improve the competitiveness of composite bridges, as a consistent next step not only the steel girders, but also the concrete deck needs to be pre-fabricated, which results in a reduction of the number of site operations and a substantial saving in the construction time and a shorter disruption of traffic flow respectively. Although already successfully built in France and Sweden, the critical details (mainly of the slab and its connections) need to be identified, proper solutions need to be found and the applicability, durability and sustainability of this composite bridge type must be proven in order to make the concept cost efficient. The major objectives of the ELEM project are to improve the overall competitiveness of this upcoming composite bridge type. Furthermore matters of safety during construction will be highlighted. The research aims at making bridges with pre-fabricated deck elements competitive not only for short and medium spans but also for multi spans, which will lead to an increase in the use of steel in bridges. The project includes both intensive theoretical and experimental investigations on steel components, concrete slab and connections, including the testing and monitoring of a composite bridge constructed with prefabricated deck elements.
Studies and reports
doi:10.2777/79809
