Cyclic Tests on External RC Beam-Column Joints

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Cyclic Tests on External RC Beam-Column Joints: Role of Seismic Design Level and Axial Load Value on the Ultimate Capacity ANGELO MASI, GIUSEPPE SANTARSIERO, and DOMENICO NIGRO JOURNAL OF EARTHQUAKE ENGINEERING, 17:110–136, 2013

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JOURNAL OF EARTHQUAKE ENGINEERING, 17:110–136, 2013

CYCLIC TESTS ON EXTERNAL RC BEAM-COLUMN JOINTS: ROLE OF SEISMIC DESIGN LEVEL AND AXIAL LOAD VALUE ON THE ULTIMATE CAPACITY Angelo Masi1, Giuseppe Santarsiero2, Domenico Nigro3 1 DiSGG, Università della Basilicata, Potenza, Italy, [email protected] 2 DiSGG, Università della Basilicata, Potenza, Italy, [email protected] 3 DiSGG, Università della Basilicata, Potenza, Italy, [email protected]

ABSTRACT A wide experimental program on beam-column RC joints carried out in the framework of the DPCReluis Project (DPC: Department of Civil Protection, Reluis: Network of University Laboratories of Earthquake Engineering) is presented. All the experimental tests were performed at the Laboratory of Structures of the University of Basilicata, Potenza (Italy). The main objective of the experimental campaign is to study and compare the post-elastic behaviour of beam-column joints with different earthquake resistant design levels, indicating the role of some structural parameters such as the axial load value acting on the column, the beam dimensions, and the steel type, on the joint performances and failure mechanism. The analyses have mainly been devoted to improving the assessment procedures regarding existing buildings but also to verifying the prediction capability of the capacity models relevant to beam-column joints contained in literature and in the new seismic codes. Following a short description of the experimental methodologies used in other campaigns, the experimental program is presented, providing a detailed description of the specimens and of the testing set-up. This is followed by a report of the main results of the cyclic tests performed on the beam-column specimens which highlight the role played by axial load and seismic design level in determining the failure mechanism and the global response of the joints.

KEYWORDS Reinforced concrete, existing buildings, beam-column joints, joint panel, cyclic test, collapse mechanism. 1 INTRODUCTION The experimental programs on Reinforced Concrete (RC) beam-column joints are generally designed to investigate the role of some design parameters, such as earthquake resistant design (ERD) level, detailing of reinforcing bars, type of bars (smooth or deformed), magnitude of the axial load acting on the column. More recent tests are also specifically devoted to studying the intersection region between beam and column members, i.e. the joint panel. Design procedures of new buildings and assessment procedures of existing ones have both generally focused on structural elements such as beams and columns and paid little attention to the intersection region. However, due to its aspect ratio, it is worth noting that the joint panel is subject to shear action several times higher than that of the framing columns. For this reason, the panel can suffer brittle failure and

threaten the capacity of the entire structure, thus reducing the benefits of an effective design of beam and column members. Indeed, past earthquakes (e.g. Abruzzo 2009) have shown that joint performances have a very important influence on the strength and the overall stability of RC framed structures. Many experimental tests on beam-column joints concentrate on buildings designed to only resist gravity loads (Hakuto at al., 2000), due to their vulnerability and wide presence in many earthquake-prone countries. The main objective is to establish reliable methods for assessing their resistance and deformability and, as a consequence, to develop new retrofit techniques. Other tests have been carried out on joints belonging to seismically designed structures with the objective of verifying the influence of some detailed solutions on the behaviour and collapse mechanism. As an example, some studies have highlighted the influence of the anchorage type within the joint adopted for the longitudinal beam bars (Park, 2002; Calvi et al., 2002; Kusuhara & Shiohara, 2008). The presence of the axial load on the column was not considered in some experimental programs (e.g. Park, 2002). However, its presence appears important to adequately reproduce the real working conditions of joints. Beyond the obvious consequences on the capacity of column members, this could have a remarkable impact on joint performances deriving from the shear strength of the joint panel, where the stresses are directly dependent on the value of the axial load acting on the column. In fact, axial force has a large influence on the shear strength of the joint as recognized in the capacity models given in the literature (e.g. Paulay & Priestley, 1992), and as confirmed by the present experimental study. In most cases, the applied axial load (normalized with respect to fc.b.h, where fc is the cylindrical concrete strength, b and h are the dimensions of the column section) is around 0.10-0.15, and in some cases reaches values around 0.30-0.40. The application of high values of the normalized axial load, especially in the case of full scale specimens, requires the application of large forces. However, while proper test apparatus is necessary to do this, it is sometimes not available. In some experimental tests axial load values varying with the imposed drift have been adopted, in order to more accurately reproduce the actual conditions under seismic actions. Indeed, for increasing drift values, and depending on the overall structural configuration (number of storeys, and then overturning moment), axial load on the columns can change considerably producing effects on their ductility and on the shear strength of the joint panel. Another factor investigated is the role of the amount of transverse reinforcement, if any, in the joint panel. These studies (e.g. Shiohara, 2001) permitted the development of models to calculate the shear capacity of the joint panel with and without transverse reinforcement, recently included in seismic codes (CEN, 2004b; 2005). Joint specimens subject to more frequent tests have cruciform shape (internal joints, e.g. Shin & LaFave, 2004) or T shape (external joints, e.g. Pantelides et al., 2002), while L shaped joints, representative of roof elements, are less frequently investigated (Calvi et al., 2002). The joints tested are frequently made in full scale, because this allows the creation of specimens utilizing exactly the same techniques and the workers who normally operate in the execution of real buildings, thus avoiding the use of scaling factors in both design and interpretation of results. However, testing joints in full scale can be difficult or beyond the capacities of the testing machine, e.g. in requiring that high values of axial load on the column are applied. For this reason, in some cases reduced scale specimens are investigated (e.g. Kusuhara et al., 2004). The loading schemes found in literature mainly differ in the point of application of the cyclic force. In the first layout the acting force is applied to the beam while the column is restrained at the ends (Hwang et al., 2004, Pantelides et al., 2002, Murty et al., 2003, Clyde et al., 2000). In a second, more realistic scheme the application of the seismic force is at the top of the column, restraining the beam end to have only horizontal displacements (e.g. Calvi et al. 2002; Shin & LaFave, 2004; Park, 2002; Braga et al., 2001). This second scheme enables a more successful achievement of inter-storey drift values at different damage states, thus facilitating the use of test results to evaluate the performances RC frames can actually experience. In fact, drift is generally considered to be an

effective indicator of global damage on RC buildings as it correlates well with their structural and non-structural damage (Naeim, 2001; Priestley et al., 2007; Masi et al., 2011). Almost all the experimental programs on RC beam-column joints found in literature were carried out by applying cyclic loading condition with low frequencies (quasi static). Cyclic loading is mandatory in order to investigate the strength degradation due to the repeated loads of seismic conditions. For this reason, even though the test can be either displacement or force controlled, the former loading mode appears more appropriate in following the specimen degrading behaviour. 2 THE EXPERIMENTAL PROGRAM The research project DPC-Reluis, funded by the Department of Civil Protection, DPC, was carried out by the Reluis consortium (Network of University Laboratories of Earthquake Engineering, www.reluis.it), in which the University of Basilicata is involved. The project had ten research Lines, where Line 2 was devoted to the “Assessment and reduction of seismic vulnerability of existing RC buildings”. In turn, each research Line was subdivided into several tasks. Task 7 in research Line 2 was named “Joints” and specifically devoted to the study of the behaviour and strengthening of RC beam-column joints. A wide experimental program has been planned, set up and partially performed within the framework of this research project, in the Laboratory of Structures of the University of Basilicata. Globally, cyclic tests on 26 beam-column joint specimens are foreseen. The main objective of the experimental program is to investigate the behaviour and the collapse mechanism of joints as a function of some geometric and mechanical factors, including the Earthquake Resistant Design (ERD) level, the beam type (stiff or flexible), and the axial load acting on the column. In order to achieve a better understanding of the role of the above factors, other characteristics (for example the concrete strength) were kept constant in all the specimens. The total number of specimens was organized in 13 pairs, each pair having identical characteristics and subjected to the same test modality. In table 1 the main characteristics of each pair are summarized. The specimens were built in July 2006, while the tests began several months later. Overall, 10 experimental tests have been carried out up to date. The specimens are representative of one-way external joints belonging to the first storey of an internal frame of a four storey RC building (reference structure). The 13 pairs can be subdivided into two main groups on the basis of their ERD level. The first group is representative of joints extracted by RC structures designed for gravity loads only, typically present in the European and Italian residential building stock, already studied by means of numerical models (Masi, 2003; Masi et al., 2011). The other group of joints is relevant to a similar structure but designed to resist earthquake loads according to the seismic code OPCM 3274 (PCM, 2003) in effect in Italy when the specimens were designed. It is worth noting that design in accordance with the OPCM 3274 is substantially consistent with the current Italian code (NTC08, 2008) as well as with the Eurocode 8 (CEN, 2004b). More specifically, the characteristics considered in the investigation are as follows: 1. ERD level of the reference structure: design for gravity loads only (No Earthquake Resistant Design, NE) matching the codes in force in Italy in the ’70s (D.M. 1972), earthquake design (E) considering both medium seismicity zone (Zone 2, ag = 0.25g) and very low seismicity zone (Zone 4, ag = 0.05g), where ag is the design ground acceleration at the ultimate limit state on type A ground (CEN, 2004b). 2. Beam type: Stiff Beams having 300x500 mm cross section (RB), and Flexible Beams having 600x240 mm cross section (FB). 3. Type of reinforcing steel: deformed bars with two different levels of ductility, that is type D steel having high ductility and type S steel with lower ductility and higher strength.

4. Amount of axial stress on the column: low (ν = 0.15), and high (ν = 0.3). With respect to the considered beam types, it is worth noting that flexible (sometimes defined flat or wide) beams are frequently present in the European building stock, as discussed in Masi (2003) as well as in other studies (e.g. Benavent-Climent et al., 2009). For this reason, testing joints with flexible beams is a matter of great concern, even though few tests on such specimens have been performed as yet. In order to facilitate the comparison between the test results, other geometrical and mechanical properties have been kept constant. Specifically, all specimens have: (i) columns with cross section dimensions 300x300 mm; (ii) concrete with an average compression strength fc around 20 MPa; (iii) steel of B450C class (steel type D, characteristic value of the nominal yield strength fyk= 450 MPa) with improved adherence (deformed bars), with the exception of two specimens (pair n. 13 in table 1), reinforced with steel type S having higher strength and lower ductility. In the seismically designed specimens (E), design forces are relevant to two different seismic zones in accordance with the previous Italian seismic classification, that is zone 4 (very low seismicity) and zone 2 (medium seismicity). In both cases a soil type A (rock or other rock-like geological formation) was considered, assuming a behaviour factor q equal to 4.095 in line with the adopted code provisions for the structural type under analysis, and following the code requirements related to a low ductility class (CD B). For low ductility structures the code does not prescribe any capacity design provisions but only an implicit value of the local ductility of RC members, achieved through reinforcement details and minimum amounts of steel. To ensure that the specimens resemble as closely as possible real building structures, the reinforcement details were designed using a software commonly adopted in design practice in Italy. Finally, with regard to axial load, the label at the relevant row in table 1 identifies the adimensional value of the force to be applied on the column during the tests. NL indicates low axial load value, equal to 15% of the ultimate compression load, that is ν = N / (bh fcm) = 0.15 (where N is the axial load, and fcm is the mean value of the concrete compression strength), while NH refers to the value ν = 0.3. As could be expected, low amounts of transverse reinforcement are present in the joints designed to gravity loads only (Figs. 1a and 1b). In particular, no hoops are placed in the joint panel and a constant spacing is maintained throughout both beam and column members. Also, the amount of longitudinal steel reinforcement is rather low, with (As/bh) values in the range 0.30-0.40%. Figs. 2 and 3 display geometry and reinforcement detailing of the joints seismically designed for the Z4 and Z2 seismic zones, respectively. As can be seen, they have identical transverse reinforcement both in beams and columns, remarkably higher than in the NE joints. The amount of longitudinal reinforcement differs only in the beam, being greater in the Z2 joint with flexible beam (1.07% against approximately 0.75% in the other specimens). This similarity, despite the difference in seismic design actions, is due to the seismic rules which prescribe rather high minimum amounts of reinforcement, irrespective of the seismic zone at hand. The reinforcement cage was prepared by workers normally engaged in the construction of building structures. Thus, no special care was adopted in constructing the specimens, which possibly have the typical defects, if any, of real structures. For this reason, before each test, a survey was performed by means of a pacometer to locate the actual arrangement of the reinforcement cage. The survey revealed that the as-built arrangement of the bars was rather different from the designed one, especially in the longitudinal bar position along the member height, therefore in the concrete cover dimensions. As an example, in the beam of the RB joints (h=500mm), which are the joints which have been tested as yet, the cover thickness of the upper longitudinal bars was on average equal to 80-90 mm, while that of the bottom bars was close to the designed one, that is about 20-30 mm. The large difference in the cover thickness was caused by an inadequate preparation of the reinforcement cage, meaning that during concrete casting the upper bars were pushed downward. Another difference with design provisions was the irregular spacing of the transverse reinforcement,

both in the beam and in the columns. All differences between designed and as-built specimens have been accounted for in the analysis of test results. During construction, each joint was provided with a steel plate 5 mm thick prior to concrete casting in order to adequately arrange the threaded bolts to link the specimen to the test apparatus. The plates were placed at the ends of the beam and columns. Contrary to other experimental tests (e.g. Braga et al., 2001), where concrete casting was made by putting the formwork on a horizontal plane, specimens were made by vertical casting, as in the real constructions. In this way, differences in concrete strength along the height of the columns can occur because of the segregation effects of the aggregates and capillary rise of water. Casting of each element of the sub-assemblage was made without time delay and finished during a continuous operation, and so interfaces are not cold joints. Each specimen was cast in a separate formwork using vertical wooden layers, as can be seen in Fig. 4. An outline of the specimens during the experimental program is displayed in Fig. 4. Figs. 4a and Fig. 4b show the joints with stiff and flexible beams, respectively. Before the execution of the cyclic tests an experimental campaign on the constituent materials was performed to characterize them from a mechanical point of view. To achieve this about 60 concrete cube specimens were prepared, during the joint casting. Before starting the tests on the joints, several concrete cubes were subjected to compression tests in order to estimate their strength, achieving a mean value equal to fc=21.5 MPa. Further tests on cubes were performed during the experimental program to detect possible concrete strength variations. Specifically, the first test (T1) was executed 9 months after joint casting when concrete mechanical properties had rather constant values so that experimental results from tests executed in different times on identical specimens could be better compared. Tests T2-T10 were executed in the later 15 months and tests on concrete cubes provided the same mean strength as that of the initial tests. Following Italian standards (NTC08, 2008) only compression tests on 150x150 mm concrete cubes were carried out. Concrete specimens to make possible tensile tests were not prepared, but tensile strength can be derived from compression strength through expressions provided in the structural codes (e.g. CEN, 2004a) and in the literature. In order to characterize the steel type D, some bars were subjected to tensile tests providing the results in terms of yielding strength fy, tensile strength ft, and ultimate strain εu reported in table 2. They are consistent with the type of steel used, B450C, in line with the current Italian structural code (NTC08, 2008) and corresponding to hot rolled steel of class C according to the Eurocode 2 (CEN, 2004a). Other tensile tests were made on the steel type S giving the results reported in table 3. In Fig. 5 typical stress-strain curves (specifically relevant to a 16 mm diameter bar) for steel type D and S are displayed. 3 TEST SET-UP Experimental tests were carried out by applying the horizontal load at the top of the column, as shown in Fig. 6 where the test apparatus is displayed. This makes it possible to directly correlate the measured displacements of the joint to the inter-storey drift of an entire frame (Pampanin et al., 2002). The axial load on the column is kept constantly equal to the value caused by gravity loads (static value). The system for applying the vertical load is able to rotate following the column, so that the load direction remains parallel to the column axis without causing P-∆ effects. Load application was cyclic quasi static with displacement control thus permitting an adequate correlation with the stiffness and strength degradation of the specimens. The axial (vertical) force was applied by a standard hydraulic jack with 1500 kN capacity, located at the top of the upper column. The jack was contrasted by a steel plate connected by four steel tie rods to a plate placed against the base of the lower column. This plate, in turn, was linked to the reaction floor by means of a pin connection so as to reproduce the same bending stress distribution

in the members of the joint specimen that can be observed when they are part of a framed structure subjected to seismic actions (bending moment with the zero point at about half inter-storey height). The beam was restrained by a vertical steel rod formed by two coupled C profiles. If second order effects are neglected, it can be considered that the restrain is a pin roller. The horizontal displacements were imposed at the end of the upper column with an MTS actuator having 490 kN push and 290 kN pull capacity. The horizontal load was imposed at a height such that the distance from the lower hinge is exactly equal to 300 cm, which is the desired inter-storey height. The test apparatus is connected to a reaction RC wall and to a RC floor both of which are 1.6 m thick, with holes purposely placed at constant spacing to allow the connection of the actuators and of the whole test system. Fig. 7 shows a picture of the test set-up before the execution of the first test. The instrumentation consists of load cells to measure applied forces and reactions, and displacement transducers (LVDTs) to measure deformations and displacements. The load cells were used to measure the axial load applied to the column, the beam reaction (consequently the beam shear), and the horizontal load applied by the actuator. The deformations of the joint panel were detected through 8 LVDT transducers (N1-N8 in Fig. 8). Furthermore, 8 LVDTs of the same type were applied to the beam (T1-T4 in Fig. 8) and to the columns (P1-P4 in Fig. 8) near to their intersection. The LVTDs used to measure the deformations of the joint panel enabled the determination of its contribution to the total drift of the specimen. Finally, 2 wire transducers were arranged for the measurement of the absolute horizontal displacements at the top of the column and at the beam free end. The test program foresees the execution of three cycles for each drift amplitude whose value increased until reaching a state either of severe damage or of near collapse, depending on the future use of the specimen, on whether it had to be repaired-strengthened or not after the first test, and then tested again. Three cycles for each amplitude are carried out because, as observed in other experimental programs, they are sufficient to evaluate the cyclic degradation of the specimen performances at constant amplitude values. The choice of the drift amplitudes was made by paying special attention to the expected cracking and yielding drift values that are around 0.5% and 1%, respectively. For this reason, the drift increasing step was smaller when the drift was below 1.5% (a step of 0.25% was chosen) and larger (0.5%) when the drift exceeded 1.5%. Thus it has been possible to better follow the cracking and yielding phenomena which occurred during the tests. It has to be noted that a drift limit equal to 0.5% is considered when verifying the damage limitation in RC buildings having non-structural elements of brittle materials attached to the structure (NTC08, 2008; CEN, 2004b). All tests considered in the present paper were performed until reaching a state of near collapse, then the maximum value of the drift was specifically chosen during each test by observing the evolution of the damage state compatible with the safety conditions of the entire test apparatus. The rate of application of the displacement is constant and equal to 4 mm/s. The whole loading history is shown in Fig. 9. 4 TEST RESULTS 10 joints with stiff beams have been tested so far. In table 4 the main results are listed in terms of characteristics of the joints (ERD and steel type), axial force value ν, failure type, and occurrence of the tensile failure in the beam longitudinal bars. The maximum force Fmax, the corresponding drift d(Fmax) and the ultimate drift du (corresponding to the ultimate chord rotation) values are also reported. The maximum force Fmax is chosen as the first peak value after yielding, while the ultimate drift du is conventionally assumed as the drift value where a force decay equal to 20% of the Fmax is observed, according to Panagiotakos and Fardis (2001). Finally, the last column of table 4 reports the drift value, db, at which the buckling of the beam bottom rebar, if any, occurred and

was visually detected once the concrete cover spalled. Additional considerations about the damage mechanisms that affected the joints are also made by referring to the principal stresses in the joint panel. The principal tension and compression stresses are evaluated with respect to the axial load N applied to the column and to the horizontal shear Vjhd calculated through equation (1) (CEN, 2005): V jhd = γ Rd ⋅ As1 ⋅ f yd − VC (1) where As1 is the area of the beam top reinforcement, VC is the column shear force (chosen as the peak value), γRd is a factor to account for over-strength due to steel strain-hardening (in this case γRd is assumed equal to 1.0 given that the mean value of the yielding steel strength has been used in equation (1)). In the following sections test results are compared in terms of observed failure mechanism and ultimate capacity. In section 4.2, results have been grouped in three groups of identical specimens, which have the same ERD level but tested under different axial load values, in order to evaluate and compare the role of axial load on joint performances. Specifically, the three groups are as follows: (i) T1, T6 and T7 tests (NE joints); (ii) T4 and T8 (Z4 joints); (iii) T2, T3, T5, T9 and T10 tests (Z2 joints). With respect to Z2 joints, it is worth remembering that the tests T9 and T10 were carried out on specimens reinforced with steel type S, that is steel having greater strength (+20% in terms of fy) and lower ductility (-80% in terms of εu) with respect to steel type D used in the other specimens. 4.1

Types of failure mechanism

Two types of failure mechanism have been observed: (i) more frequently, a purely flexural mechanism involving essentially the beam member (B mechanism), with wide sub-vertical cracks at the beam-column interface (Fig. 10a); (ii) in a few cases, a mixed mechanism in which the flexural damage to the beam was accompanied by a diagonal cracking in the joint panel (B+J mechanism, Fig. 11). It is worth noting that, in the (B+J) mechanism, cracking in the beam at the column interface was firstly observed, followed by cracking in the joint panel. These two strength degradation effects together lead the joint towards collapse. No yielding or failure in the hoops was generally observed. To gain a better understanding of Fig. 11, it should be specified that the horizontal and vertical lines represent the position of the reinforcement bars. These lines were drawn after a pacometer survey to detect the actual as-built bars’ position in order to avoid them during the installation of the measurement instruments before testing. The irregular lines display the largest cracks which occurred during the test. In Fig. 10b the failure of the bottom longitudinal reinforcing bars in the beam is shown. This failure has always been observed when mechanism type B has occurred. In fact, the wide sub-vertical crack in the beam did not completely close at the inversion of the bending moment and the beam longitudinal bars were subjected to buckling causing high local deformations leading to oligo-cyclic fatigue failure. All the failures of the beam bars occurred at the bottom because of the lower value of the concrete cover with respect to the top bars, and thus of the greater value of the arm with which the bottom bars counteracted the external bending moment. As already stated, in real RC buildings such a defect can be caused during concrete casting which pushes the longitudinal bars downward. With respect to the role of transverse reinforcement in the beam, no provisions were adopted in non seismic joints regarding the distance of the first stirrup from the beam end section, in line with the adopted code in force in non seismic zones. Contrarily, the recent Italian seismic code and the Eurocode 8, regulate the design of seismic joints by placing the first stirrup within the first 50 mm from the beam end section, as can be seen in figs. 2 and 3. However, it is worth remembering that the specimens were built by workers normally engaged in the construction of building structures, and that no special care was adopted in constructing the specimens. As a consequence, the distance of the first stirrup from the beam end section was generally larger than 50 mm. Specifically, in most

of the seismic joints it was between 60 and 70 mm. However, this cannot explain the bar buckling because in most cases this phenomenon occurred between the first and second stirrup.

4.2

Evaluation of the ultimate capacity

The tests T1, T6 and T7 on the non seismic joints (NE) were carried out with high axial force (ν=0.30) on the specimen T6 and low (ν=0.15) on the specimens T1 and T7. In Fig. 12 the sheardrift behaviour (the bold line is the envelope curve of the maximum F values at each cycle amplitude) and the damage state at the end of the test are shown. The ultimate drift values du of the joints are around 3.0%, being greater for the T7 specimen. The ultimate strength Fu is in the range 19-21 kN achieved with a drift value d(Fmax) around 0.45%. In all joints the damage mechanism only involved the beam that finally collapsed because of the fatigue failure of the bottom bars. Looking at the final damage state of T1, T6 and T7 specimens (fig. 12), it can be noted that T6 has two vertical cracks in the beam, while a single crack can be observed in T1 and T7. Even though this difference could have reduced strain concentration in the bottom beam rebar of T6, it did not affect ultimate deformation capacity that was mainly governed by the buckling of the bottom beam rebar, in turn started by the deterioration of the concrete cover. As it can be noted, the shape of hysteresis loops in the positive and in the negative loading directions are quite different. This is mainly due to the early deterioration of the bottom concrete cover occurred in the beam member with respect to the top concrete cover because the thickness of the latter was larger than the former, as already stated. Moreover, the bottom beam rebars were affected by buckling phenomena that further reduced the compression force in the bottom part of the beam section. The tests T4 and T8 on the Z4 seismic joints were carried out with the same value of the axial force (ν=0.30). In Fig. 13 the shear-drift behaviour and the damage state at the end of the test are shown. As can be seen in table 4 the joints exhibited similar values of the ultimate force of around 43 kN. The du values in these joints are practically the same, that is equal to 3.4%. In this case the collapse was also always caused by the failure of the bottom longitudinal bars of the beam. Despite having the same du values, it has to be noted that the test T4 stopped at a drift equal to 4.5%, while test T8 stopped at 3.5%. Both tests, as mentioned above, stopped because of the bottom beam rebars’ failure. This difference can be explained by looking at the db (buckling drift) values reported in table 4. In fact, for joint T8 the beam rebars’ buckling occurred at a drift equal to 2.5%, while for joint T4 at 3.0%. This early buckling speeded up the oligo-cyclic fatigue failure of the rebars of T8 specimen. The reason for the early buckling in T8 can be found in the as-built cage arrangement: in T8 concrete cover of bottom beam rebars was about 30 mm, while in joint T4 it was in the range 45-50 mm. The tests T2, T3, T5, T9 and T10 on the Z2 seismic joints were carried out with different values of the axial force. Specimens involved in tests T9 and T10 had S steel type. With respect to the Z2 joints with D steel type, Fig. 14 shows the shear-drift behaviour and the damage state at the end of the test. As can be seen in table 4, they exhibited Fmax values around 40 kN, surprisingly lower than the values obtained for Z4 joints. This can be ascribed to code provisions on the minimum amount of longitudinal reinforcement, prescribed indiscriminately in the seismic zone, thus determining very similar reinforcement amounts in Z4 and Z2 joints. The normal variability in material properties and in reinforcement arrangement details allowed the strength values of Z4 specimens to be slightly higher than those obtained for Z2 specimens. Furthermore, identical Z2 joints exhibited different behaviour. In fact, large differences of the du values can be observed between specimen T3 and the other two specimens. The T2 joint, due to local construction defects suffered an early buckling of one of the bottom beam rebars that induced an unexpected torsion effect on the beam during the test. As a consequence, a relevant reduction of the ultimate deformation capacity was found.

In the test T5 a mixed type mechanism (B+J) was observed resulting in a lower deformation capacity as a consequence of the contribution to strength decay caused by the joint panel cracking. Cracking initially occurred at beam-column interface, while for drift values over 2% diagonal cracking in the joint panel also occurred, probably due to concrete tension failure. This latter speeded up the strength drop visible in figure 14 (bottom) relevant to the Force-Drift envelope of joint T5. No fracture in the hoops was observed. Indeed, the only difference between T5 and the other two tests was the axial load value, that is equal to ν=0.3 for T2-T3 and ν=0.15 for T5. As can be seen in table 5, the principal tension stress pt in T5 is over 50% larger than in T3 (1.60 vs 1.04 MPa) and this can explain joint panel damage in T5. Cracking in the joint panel increased the loss of adherence between the longitudinal beam bars and the concrete of the joint panel, quickly reducing the flexural strength of the beam and, eventually, conditioning the response of the entire assemblage. Moreover, the maximum value of the shear force during the test T5 was attained at a higher drift value d(Fmax) (1.16% vs 0.99% of the T3 joint and 0.89% of the T2 joint) demonstrating a higher deformability of the assemblage, probably due to early slipping phenomena of the longitudinal beam bars. Contrarily to the behaviour of column members under seismic actions, which show reduction of the deformation capacity with increasing axial force (e.g. Park and Paulay, 1975), the reduction of the axial load caused a series of mechanisms that reduced the ultimate deformation capacity of the assemblage. However, it has to be noted that the range of variation of axial loads adopted in the cases studied did not affect the strength hierarchy between beam and column. Finally, in Fig. 15 the shear-drift behaviour and the damage state at the end of the test of the specimens T9 and T10 are shown. They showed higher values of Fmax because of the different type of reinforcing steel (type S) which has a yielding stress about 20% higher than the steel used for all the other specimens (type D). Both specimens exhibited the same mixed failure mechanism. Indeed, the higher steel strength increased the shear demand Vjhd in the joint panel (see eq. (1)). Although the T9 specimen was tested under a high axial force (ν=0.3) joint panel cracking occurred, differently from the T2-T3 specimens reinforced with steel type D. In fact, table 5 shows how principal tension stress in T9 is 40% higher than that evaluated in joints T2 and T3. However, an effect of the axial load value is visible by noting the different values of d(Fmax) between joints T9 and T10. The T9 specimen tested under a higher value of the axial load has d(Fmax)=1.50%, less than the value obtained for the T10 specimen (d(Fmax)=1.95%) tested under low axial load (ν=0.15). In this case results suggest that, although the higher value of the axial load could not avoid cracking of the joint panel, it allowed a better adherence condition of beam reinforcement as shown by the lower value of d(Fmax). Analysing eq. (1), the absence of cracking in the joint panel of NE specimens, despite the lack of transverse reinforcement, can be explained: the term As1 is very small with respect to seismic joints (2.26 cm2 vs 5.32-6.03 cm2), causing a lower shear demand which is less than the joint panel strength as shown by the low values of the principal stresses reported in table 5. They are several times less than those computed in seismic joints where a larger amount of steel area in the beam (As1) is placed. Indeed, the absence of joint panel cracking is a peculiar characteristic of the non seismic joints in this experimental program, and is a consequence of the low amount of longitudinal reinforcement in the beam. However, the behavior of seismic joints is globally more favourable because of their higher strength combined with a good ultimate deformation capacity, and their capacity to adequately support vertical loads even after the joint panel cracking. It has to be noted that the ultimate deformation capacity du reported in table 4 is always the conventional value assigned following the method proposed in Panagiotakos and Fardis (2001). The maximum drift values applied during the tests are larger than such conventional values, although tests were not performed up to “real” collapse of the specimens, i.e. total loss of their load bearing capacity. Maximum values correspond to the failure of one or more beam rebars (for joints affected by B failure type) or to a damage state beyond which safety of operators and test apparatus could become critical (for joints affected by B+J failure type).

Results highlight that the conventional value of ultimate drift corresponding to a strength drop of 20% is a conservative choice since all the joints survived to this drift value. Given the failure mechanisms observed for the tested joints, it is interesting to check the effectiveness of EC8 (CEN, 2004b) expressions in predicting the shear capacity of the joint panel. The EC8 formulation for predicting the joint shear capacity has two separate steps. Firstly, there is an expression to evaluate the compression capacity of the strut that can be recognized in the joint panel under seismic actions and, secondly, an expression to compute the tensile strength of the joint in order to avoid diagonal cracking is provided. The horizontal shear demand Vjhd should not exceed a value that could cause the compression failure of the joint: V jhd ≤ η f cd 1 −

νd b j ⋅ h jc η

(2)

where η = 0.60 (1-fck/250) for interior joints and η = 0.48 (1-fck/250) for exterior joints, practically meaning that the strength of exterior joints is 0.8 (0.48/0.60) times that of interior joints (assuming the same joint materials and detailing); νd is the normalised axial force in the column above the joint, fck is given in MPa, hjc is the distance between the extreme layers of column reinforcement, bj is the effective width of the joint. Furthermore, EC8 provides an expression to design the joint transverse reinforcement (left hand term in the following expression (3)) needed to avoid the diagonal cracking caused by the achievement of the concrete tensile strength fctd, as follows: 2 Ash ⋅ f ywd (V jhd / b j ⋅ h jc ) ≥ − f ctd (3) b j ⋅ h jw f ctd + ν d f cd where, Ash is the total area of the horizontal hoops in the joint, Vjhd is the horizontal joint shear demand, hjw is the distance between top and bottom reinforcement of the beam. Therefore, the code recognizes two possible failure mechanisms of the joint panel. Using the eqs. (2) and (3) with the equality sign, the value of Vjhd can be computed achieving, respectively, the compression shear strength (Vjc) and the tension shear strength (Vjt). Given that the above formulation is related to joints consistent with the code provisions, expressions 1-3 have been applied for all the seismic joints (T2, T3, T4, T5, T8, T9, T10) in order to verify the predictive capacity of the code capacity models and, for the joints whose panel failed, to identify the cause of the failure (tension or compression). It is important to underline that expressions (2) – (3) have been used to calculate the actual capacity of the joints and not as conventional safety verifications, therefore no safety factors have been adopted. More specifically, design and characteristic compression strength of concrete have been assumed equal to the mean value, that is fcd = fck = fcm = 21.5 MPa. For the tension strength of concrete it has been assumed fctd= fctm = 0.3(fcm)2/3. The steel design strength has been assumed as follows: • fyd = fydw=fy= 480 MPa for steel type D. • fyd = fydw=fy= 580 MPa for steel type S. Adopting the mean value for the steel strength as mentioned above, the over-strength factor has been assumed equal to unity: γRd = 1. Results are displayed in Fig.16 from which it can be noted that: • in all the joints where the panel did not fail (T2, T3, T4, T8) horizontal shear demand was lower than shear strength for both tension (Vjt) and compression (Vjc); • In T10 joint the shear demand was higher than tensile shear strength, then collapse should be ascribed to a tensile concrete failure; • In T9 joint the shear demand was higher than compression shear strength, then collapse should be attributed to a compression failure of concrete;

•

T5 joint should not have failed according to the code formulation, although experimental evidence shows the contrary. With respect to T9 and T10 joints, really there was little difference in the observed failure mechanism. Therefore, considering that the difference between the predicted shear strength and the actual shear demand are quite small, there is the possibility that the two specimens failed basically because of the same mechanism. Thus, the EC8 formulation shows a good estimation capacity regarding the occurrence of joint panel failure, with the exception of the wrong prediction concerning the T5 joint. Such a discrepancy could derive from the variation of steel strength among different joints. For example, steel hardening at high deformations can increase the shear demand Vjhd thus exceeding Vjt, able to better match the experimental and code results for the T5 joint. In accordance with almost all experimental studies found in the literature (e.g. Hakuto et al 2000, Calvi et. Al. 2002, Park 2002), specimens without floor slabs were constructed and tested. In the cases studied the floor slab, as permitted by the Italian structural code, was made up of one-way joists with non-structural clay elements placed between the joists and a concrete topping 40mm thick. Therefore, the addition of a slab plate 40 mm thick would increase the positive resisting moment of the beam by around 15% for the Z2 joints (i.e. those provided with the larger amount of beam reinforcement). Such an increase would not have changed the strength hierarchy, also for the non seismic joints having a lower amount of beam reinforcement. The presence of the floor slab and the possible presence of orthogonal beams could have a real influence on the performances of beam-column joints particularly with respect to the damage and collapse mechanism of the joint panel. Thus its inclusion in future research programs should be considered, even though it could lead to significant complications in the execution of the tests. 4.3

Joint panel deformations

Evaluating the amount of joint panel deformability can be important since the models usually adopted assume joint panel rigidity, and thus do not account for its contribution to the deformability of the whole structure. The instruments installed to measure the local deformations of the joint panel permitted the evaluation of its contribution to the total drift of the assemblages. Specifically, the shear distortion of the joint panel was measured by means of 8 LVDTs, as can be seen in figure 8. Unfortunately, it was not always possible to record the shear distortion values throughout the test, as damage to joint panels, in some cases, caused the detachment of the instruments at the higher drift values. To obtain a better analysis of the contribution of joint panel deformability, the global drift measured during the tests has been subdivided into three components ascribed to the upper column, the lower column and the joint panel deformation, respectively. The contribution of the beam was not explicitly specified, but this source of deformability has been incorporated in the drift components of the upper and lower columns. The horizontal displacement of the top of the lower column, DLC, has been calculated as the difference between the horizontal displacement of the beam end, Dbeam, and the value DT3 + DT4 measured by the LVDTs T3 and T4 (see fig. 8): DLC = Dbeam- (DT3+ DT4). The joint panel contribution has been evaluated from the horizontal component of the displacement ∆N1 measured by the diagonal N1 transducer (see fig. 8). Finally, the drift of the upper column has been obtained by subtracting the lower column and joint panel components from the global drift measured by the top wire transducer (see fig. 6). Test results show that joint panel deformability depends significantly on the collapse mechanisms that affect the beam-column assemblage. Figures 17 and 18 display the relationship between the contribution of each individual member (upper and lower column, joint panel) to the total drift of the assemblage, for the specimens T10 and T1, respectively. The amount of drift due to the joint panel is more evident in joint T10, that suffered a (B+J) mechanism with diagonal cracking in the panel, than in joint T1. For a total drift value equal to 3%, joint panel drift dJP is about 0.5% in the

specimen T10 (i.e. about 16% of the total drift), while dJP values around 0.1% were measured in the joint panel of the specimen T1 (i.e. about 3% of the total drift). In figure 19 results for all tested specimens are reported, displaying the joint panel contribution (%) at 3% total assemblage drift. The larger dJP values were measured in the specimens T5, T9 and T10 that suffered a (B+J) collapse mechanism, equal to 13%, 7% and 16% of the total drift, respectively. The smaller value recorded for joint T9 among these results, can be ascribed to the higher level of axial force (ν=0.3) that determined a better confining effect and thus reduced damage in the joint panel. When the joint panel is not damaged (tests T1, T2, T3, T4, T6, T7, T8), its drift contribution is small with values always under 5% of the total drift and hence joint panel deformability seems negligible in cases of collapse mechanism where the joint panel is not involved. Furthermore, also in the specimens that suffered joint panel cracking, its contribution becomes significant only for high values of the total drift, i.e. around 2% in the cases studied. 5 FINAL REMARKS After a short literature review on objectives, test modalities, apparatus and equipment of the experimental investigations on RC beam-column joints, some initial results of a wide experimental program carried out in the framework of an Italian National research project are presented in the paper. 26 joints were designed and built at the Laboratory of Structures of the University of Basilicata (Italy) organized in 13 pairs, each one having identical characteristics and general test modality. The specimens, representative of external joints belonging to the first storey of a four storey RC building, can be subdivided on the basis of the ERD level (gravity load or seismically designed), beam type (stiff or flexible), type of reinforcing steel (higher ductility or higher strength), and amount of axial stress on the column (low or high). The construction of the specimens was made paying due attention to reproduce actual construction conditions, for example casting the concrete in vertical direction and using the workmanship usually employed in the real RC constructions. 10 (all with stiff beam RB) out of the 26 joints have been tested up to now, providing some initial results on the failure mechanism and behaviour of such structural members. Firstly, specimens with the same material properties and details showed different performances, in terms of both failure mechanism and shear-drift behaviour, depending on the value of the axial force applied to the column. In fact, two main failure modes have been observed: flexural failure involving only the beam member, and mixed failure where, in addition to the beam, the joint panel was also damaged. When comparing the two mechanism types, remarkably lower deformation capacities in the specimens affected by mixed failure have been found. The failure mode appears to be dependent on the axial load value because of its influence on the shear strength of the joint panel. Specifically, joint panel cracking has been generally observed with the lower axial load values and, moreover, only in seismically designed joints. Non seismic joints, even though lacking in transverse reinforcement in panel joints, do not experience panel cracking because of the low amount of longitudinal reinforcement in beams and, consequently, of the lower values of shear demand on joint panels. In fact, under axial load and shear demand (computed through eq. (1)) non seismic joints have shown values of the principal tension stress several times lower than the ones obtained for seismic joints. With respect to seismically designed specimens, joints having identical ERD level and steel type, tested under different axial load, generally suffered different collapse mechanisms due to the different stress conditions in the joint panel. As an example, the joint T5, tested with the lower value of the axial load, showed severe damage to the panel, but not exhibited by the identical joint T3 tested with the higher value of the axial load. However, it should be noted that although seismic specimens can suffer joint panel cracking, their performances are in any case more favourable than non seismic specimens since the presence of hoops provides adequate capacity to support gravity

loads also in the event of severe earthquakes. According to EC8, the detailing of seismic joints is able to ensure the integrity of the panel also after heavy cracking of the beam-column intersection. Besides, it has to be noted that seismic joints provide an ultimate deformation capacity larger than or equal to the one exhibited by non seismic joints combined with a higher level of shear response. The effectiveness of EC8 (CEN, 2004b) expressions in predicting the shear strength of seismically designed beam-column joints has been checked demonstrating a good estimation capacity. Joint panel cracking is predicted except in the case of joint T5, where shear strength and demand are, however, comparable. Another noteworthy as well as unexpected finding is represented by the tension failure of the longitudinal bars in the beams of joints affected by a purely flexural collapse. In fact, the wide subvertical crack in the beam did not completely close at the inversion of the bending moment and the beam longitudinal bars were subjected to buckling causing high local deformations leading to oligocyclic fatigue failure. Such a result suggests that adequate attention has to be devoted to a failure mechanism which has usually been neglected, as in the current Italian structural code. In particular, it has to taken into account that the buckling of the bars that favoured their failure, generally occurs between the first and second stirrup. Therefore, further studies on the provisions currently provided in the codes concerning spacing of transverse reinforcement within the critical regions of beam members should be made. Finally, the joint panel deformations have been measured and their contribution to the total drift of the assemblages has been evaluated. Test results show that joint panel deformability depends significantly on the collapse mechanism that affects the assemblage. The larger values were measured in the specimens T5, T9 and T10 that suffered a (B+J) collapse mechanism, equal to 13%, 7% and 16% of the total drift, respectively. When the joint panel is not damaged (tests T1, T2, T3, T4, T6, T7, T8), its contribution is small with values always under 5% of the total drift and hence joint panel deformability seems negligible in cases of collapse mechanism where the joint panel is not involved. Furthermore, also in the specimens that suffered joint panel cracking, its contribution becomes significant only for high values of the total drift, i.e. around 2% in the cases studied. The above initial findings indicate that more accurate analyses should be performed and also supported by further studies and tests, in particular concerning the role of the axial load values on both local and global ductility of RC framed structures. The completion of the experimental program is currently in progress. Further results on joints with stiff beams as well as those drawn from the investigation of the peculiar performances of joints with flexible (flat) beams will be presented in future papers. ACKNOWLEDGEMENTS The work reported in the present paper has been carried out within the framework of the DPCReLUIS 2005-2008 Project, Research Line n. 2 “Assessment and reduction of the vulnerability of RC existing buildings” (Task JOINTS). The contribution of Ferriere Nord S.p.A. Gruppo PITTINI, Arcasensa SAS and F.lli De Stefano s.n.c. firms in the manufacturing of the joint specimens is gratefully acknowledged. Finally, the authors wish to express their gratitude to the three anonymous reviewers for the detailed comments and suggestions provided, which have been very helpful in improving the paper. REFERENCES Benavent-Climent, A., Cahís, X., Zahran, R. [2009] “Exterior wide beam-column connections in existing RC frames subjected to lateral earthquake loads”, Engineering Structures, 31, 14141424.

Braga, F., De Carlo, G., Corrado, G.F., Gigliotti, R., Laterza, M., Nigro, D. [2001] “Meccanismi di risposta di nodi trave-pilastro in c.a. di strutture non antisismiche” Proc. of the 10th National Conference “L’ingegneria Sismica in Italia”, Potenza-Matera (in Italian). Calvi, G.M., Magenes, G., Pampanin, S. [2002] “Relevance of beam - column joint damage and collapse in RC frame assessment” Journal of Earthquake Engineering, Vol. 6, special Issue 1, 75-100. CEN [2004a] Comitè Europeén de Normalisation EN 1992-1-1:2004 Eurocode 2: Design of concrete structures - Part 1-1: General rules and rules for buildings, 2004, Brussels. CEN [2004b] Comitè Europeén de Normalisation EN 1998-1:2004 Eurocode 8: Design of structures for earthquake resistance—Part 1: general rules, seismic actions and rules for buildings, December 2004, Brussels. CEN [2005] Comitè Europeén de Normalisation EN 1998-3:2005 Eurocode 8: Design of structures for earthquake resistance - Part 3: Assessment and retrofitting of buildings, June 2005, Brussels. Clyde, C., Pantelides, C. P., Reaveley, L. D. [2000] “Performance-Based Evaluation of Exterior Reinforced Concrete Building Joints for Seismic Excitation” PEER Report 2000/05, University of California, Berkeley. D.M. 1972, Ministry of Public Works [1972] D.M. 30/05/1972 Technical prescriptions to which the normal and pre-stressed reinforced concrete and steel construction have to undergo, (in Italian). Hakuto, S., Park, R., Tanaka, H. [2000] “Seismic load tests on interior and exterior beam-column joints with substandard reinforcing details”, ACI Structural Journal, Vol. 97, No. 1. Hwang, S., Lee, H., Wang, K. [2004] “Seismic design and detailing exterior reinforced concrete beam column joints,” 13th World Conference on Earthquake Engineering, Vancouver, B.C., Canada, August 1-6, 2004 Paper No. 397. Kusuhara, F., Shiohara, H. [2008] “Tests of R/C beam-column joints with variant boundary conditions and irregular details on anchorage of beam bars”, 14th World Conference on Earthquake Engineering, October 12-17, 2008, Beijing, China. Kusuhara, F., Akukawa, K., Shiohara, H., Otani, S. [2008] “Tests of reinforced concrete interior beam-column joint subassemblage with eccentric beams”, 13th World Conference on Earthquake Engineering, August 1-6, 2004, Vancouver, Canada. Masi, A., [2003] “Seismic vulnerability assessment of gravity load designed R/C frames,” Bulletin of Earthquake Engineering, Vol. 1, N. 3, pp. 371-395. Masi, A., Vona, M., Mucciarelli, M. [2011] “Selection of natural and synthetic accelerograms for seismic vulnerability studies on RC frames,” Journal of Structural Engineering, Vol. 137, N. 3, DOI: 10.1061/(ASCE)ST.1943-541X.209. ISSN: 0970-0137. Murty, C. V. R., Durgesh, C. R., Bajpai, K. K., Jain, S. K. [2003] “Effectiveness of Reinforcement Details in Exterior Reinforced Concrete Beam-Column Joints for Earthquake Resistance”, ACI Structural Journal Vol. 100, No. 2, pp. 149-156. Naeim, F. (ed.), [2001] The seismic design handbook, 1st Edition, Kluwer Academic Publishers, The Netherlands. NTC08 [2008] D.M. 14 gennaio 2008—Norme tecniche per le costruzioni (in Italian). Ministero delle Infrastrutture. Available at http://www.cslp.it Pampanin, S., Calvi, G.M., Moratti, M. [2002] “Seismic behaviour of R.C. beam-column joints designed for gravity loads” 12th European Conference on Earthquake Engineering, EAEE, London. Panagiotakos, T.B., Fardis, M.N. [2001] ”Deformation of reinforced concrete members at yielding and ultimate”, ACI Structural Journal, Volume 98, No. 2, 135-148. Pantelides, C.P., Hansen, J., Nadauld, J., Reaveley, L.D. [2002] “Assessment of reinforced concrete building exterior joints with substandard details”, PEER Report 2002/18, University of California, Berkeley.

Park, R., [2002] “Summary of results of simulated seismic load tests on reinforced concrete beamcolumn joints, beam and column with substandard reinforcing details”, Journal of Earthquake Engineering, Vol. 6, No.2, 147-174. Park, R., Paulay, T. [1975] Reinforced Concrete Structures, J. Wiley & Sons, New York. Paulay, T., Priestley, M. J. N. [1992] Seismic design of reinforced concrete and masonry buildings, John Wiley & Sons, New York. PCM [2003] Presidenza del Consiglio dei Ministri, “OPCM 3274 e s.m.i. - Allegato 2 Norme tecniche per il progetto, la valutazione e l’adeguamento sismico degli edifici”, G.U. 8/5/2003 (in Italian). Priestley, M.J.N., Calvi, G.M., Kowalsky, M.J. [2007] Displacement-Based Seismic Design of Structures, IUSS Press, Pavia, Italy. Shiohara, H. [2001] “A New Model For Joint Shear Failure of Reinforced Concrete Interior Beamto-Column Connections”, ASCE Journal of the Structural of Engineering, Vol. 127, No. 2, 152160. Shin, M., LaFave, M.J. [2004] “Modelling of cyclic joint shear deformation contributions in RC beam-column connections to overall frame behavior”, Structural Engineering and Mechanics, Vol. 18, No.5, 645-669.

Figure 1. Reinforcement detailing of non seismic joints (NE) with a) stiff beam (RB) and b) flexible beam (FB).

Figure 2. Reinforcement detailing of joints for seismic zone 4 (Z4) with a) stiff beam (RB), b) flexible beam (FB).

Figure 3. Reinforcement detailing of joints for seismic zone 2 (Z2) with a) stiff beam (RB), b) flexible beam (FB).

Figure 4. Outline of the specimens after casting: a) joints with stiff beam, b) joints with flexible beam.

Figure 5. Stress-strain curves of type D and type S steel for 16 mm diameter rebars.

Figure 6. Test set-up.

Figure 7. Picture of the test set-up with a specimen before testing.

Figure 8. LVDTs location on a specimen with stiff beam (RB).

Figure 9. Loading history.

Figure 10. Flexural mechanism on the beam (failure type B): a) beam crack pattern, b) failure of bottom beam bars at the end of the test (Joint T1).

Figure 11. Mixed mechanism (Joint T5).

Figure 12. Tests T1, T6 and T7 on non seismic joints (NE): shear-drift behaviour and final damage state.

Figure 13. Tests T4 and T8 on joints designed for seismic zone 4 (Z4): shear-drift behaviour and final damage state.

Figure 14. Tests T2, T3 and T5 on joints designed for seismic zone 2 (Z2) with D steel type: shear-drift behaviour and final damage state.

Figure 15. Tests T9 and T10 on joints designed for seismic zone 2 (Z2) with S steel type: shear-drift behaviour and final damage state.

Figure 16. Shear demand (Vjhd) vs shear strength (Vjc, Vjt) in the seismically designed specimens.

Figure 17. Contribution of upper and lower columns and of joint panel to the total drift of assemblage in case of (B+J) collapse mechanism (results of test T10).

Figure 18. Contribution of upper and lower columns and of joint panel to the total drift of assemblage in case of B collapse mechanism (results of test T1).

Figure 19. Joint panel contribution (%) at 3% total assemblage drift in all tested specimens.