Seismic performance of an existing bridge with scoured caisson ...

Vol.13, Suppl.1

EARTHQUAKE ENGINEERING AND ENGINEERING VIBRATION

Earthq Eng & Eng Vib (2014) 13: 151-165

August, 2014

DOI: 10.1007/s11803-014-0245-x

Seismic performance of an existing bridge with scoured caisson foundation Kuo-Chun Chang1†, Yu-Chi Sung2†, Kuang-Yen Liu3‡, Ping-Hsiung Wang3§, Zheng-Kuan Lee3‡, Lu-Sheng Lee3* and Witarto4+ 1. Department of Civil Engineering, Taiwan University (NTU), Chinese Taipei 2. Department of Civil Engineering, Taipei University of Technology (NTUT), Chinese Taipei 3. Center for Research on Earthquake Engineering (NCREE), Chinese Taipei 4. Department of Civil Engineering, The University of Houston, Houston, USA

Abstract: This paper presents in-situ seismic performance tests of a bridge before its demolition due to accumulated scouring problem. The tests were conducted on three single columns and one caisson-type foundation. The three single columns were 1.8 m in diameter, reinforced by 30-D32 longitudinal reinforcements and laterally hooped by D16 reinforcements with spacing of 20 cm. The column height is 9.54 m, 10.59 m and 10.37 m for Column P2, P3, and P4, respectively. Column P2 had no exposed foundation and was subjected to pseudo-dynamic tests with peak ground acceleration of 0.32 g first, followed by one cyclic loading test. Column P3 was the benchmark specimen with exposed length of 1.2 m on its foundation. The exposed length for Column P4 was excavated to 4 m, approximately 1/3 of the foundation length, to study the effect of the scouring problem to the column performance. Both Column P3 and Column P4 were subjected to cyclic loading tests. Based on the test results, due to the large dimension of the caisson foundation and the well graded gravel soil type that provided large lateral resistance, the seismic performance among the three columns had only minor differences. Lateral push tests were also conducted on the caisson foundation at Column P5. The caisson was 12 m long and had circular cross-sections whose diameters were 5 m in the upper portion and 4 m in the lower portion. An analytical model to simulate the test results was developed in the OpenSees platform. The analytical model comprised nonlinear flexural elements as well as nonlinear soil springs. The analytical results closely followed the experimental test results. A parametric study to predict the behavior of the bridge column with different ground motions and different levels of scouring on the foundation are also discussed. Keywords: in-situ test; seismic performance; bridge column; scouring; soil-structure interaction

1. Introduction Generally, the seismic performance of structures can be assessed by conducting either laboratory experiments or in-situ tests. While many researchers have adopted scaled-down specimens for laboratory experiments, laboratory experiments using full-scale specimens have been developed in recent years to avoid any scale-effects between laboratory samples and in-situ tests (Kazuhiko et al., 2009; Kawashima et al., 2012; Paolo et al., 2000). Several discrepancies exist between laboratory experiments and in-situ tests: (1) laboratory experiments frequently adopt foundations that are secure and not embedded in soil that may be unstable, whereas soil-structure interactions are clearly demonstrated in onCorrespondence to: Kuang-Yen Liu, Center for Research on Earthquake Engineering (NCREE), Chinese Taipei Tel: +886-2-66300586; Fax:+886-2-66300858 E-mail: [email protected] † Professor; ‡Associate Researcher; §Associate Technologist; *Technologist; +Ph.D Student Received June 20, 2014; Accepted July 28, 2014

site bridge columns; (2) the loads of the superstructure can differ between laboratory experiments and in-situ tests; laboratory experiments adopt assumptions of the axial forces on a column, while the forces applied to on-site bridges are variable; (3) laboratory experiments are subjected to size effect; therefore, conclusions (e.g., actual structural behavior of the bridges) derived from these types of experiments can be debatable; and (4) variations in construction details exist; while the quality of construction for laboratory experiments is relatively precise, the variable influence of on-site construction and environmental factors are not incorporated into the experimental conditions. Consequently, when the installation of a reaction and loading system is permitted, an in-situ test is still necessary for determining the actual behavior of the structure systems. In-situ tests on structure systems have been carried out in the past. Ko and Chen (2010) generated a pushover curve by considering foundation flexibility, and the results suggested the importance of soilstructure interactions. Costa et al. (2011) reviewed the before and after effects of reinforcing buildings with

152


bricks. Regarding bridge structures, Pantelides et al. (2001) introduced the cyclic loading test for two-column columns (Bents #5N, #6N, and #5R) that were erected at the South Temple Bridge in 1998. They investigated the lack of seismic performance in the plastic hinge and column joint regions of the bridge that was designed before 1960, and employed carbon fibers to confirm the reinforcement effects. Cyclic loading tests on bridges (Bents #4S, #5S, and #6S) situated south of the South Temple Bridge have also been conducted (Pantelides et al., 2003). Pantelides and Fitzsimmons (2012), examined the reinforcement effects and the cost implications to bridges by employing National Earthquake Hazards Reduction Program (NEHRP)-designed earthquakes. Distinctive of bridge column components, Chai and Hutchinson (2002) conducted an in-situ test on piles to investigate the effects that confining steel ratios, aboveground height, and soil density have on the bending strength and displacement ductility of piles that are embedded in non-cohesive soils. Houlsby et al. (2005) conducted a foundation vibration test by subjecting a 3 m diameter suction caisson into soft clay and employing a foundation cyclic loading test with a 1.5 m diameter suction caisson. Subsequently, the laboratory test results were compared to the normalized structural reaction (Houlsby et al., 2006). Previous researchers have focused on seattype and monolithic abutments to conduct full-scale experiments. Subsequently, they proposed a non-linear analytical model for relevant abutments with backfill (Romstad et al., 1995; Stewart et al., 2007; Lemnitzer et al., 2009 ). In summary, bridge structures are situated in natural environments; therefore, they experience changes in boundary conditions that are greater than those that can be simulated in experimental conditions in a laboratory. In addition, seismic designs for old bridges generally fail to meet the most recent specifications and requirements. Thus, to attain an ideal measure of the seismic capability of a bridge, in-situ tests are necessary. In the context of increasing concerns about a range of hazards related to bridge safety, this study investigated the effects that scouring on a gravel-based caisson foundation has on the seismic property of columns.

foundation has a diameter of 4 m and length of 12 m. The right bridge, which was built in 1961, has a total length of 259 m with a net width of 4.6 m (total width = 5.4 m). The right bridge supported by an elliptical wall column structure with the major axis of 4.4 m and minor axis of 2 m at the base. The caisson foundation, which also has an elliptical shape, originally has the major and minor axes of 9 m and 4 m, respectively. After the application of additional reinforcement, the sizes of the foundation on Column P2 to Column P6 were increased to 12 m in the major axis and 9.6 m in the minor axis. The left and right bridges were 10 m apart from centerlines with 4.5 m of the net distance between the bridge decks. Since the Niu-Dou Bridge was located at a constriction in the Nan-Yan River, the bridge foundations were exposed to intense scouring, particularly during typhoons and heavy rainfalls. Although the bridge column and foundation protection measures have been adopted in recent years, Typhoon Talim in September 2005 caused erosion of the road foundations around the A2 Abutment of the bridge, forming a 10 m long and 15 m deep hole. Furthermore, due to rapid flooding, Column P6 of the left bridge subsided by 0.77 m. Moreover, Column P6 of the right bridge was unable to withstand the scouring, resulting in downward tilt. Consequently, the deck subsided by 0.4 m and the expansion joint was translated by 0.3 m. To prevent the risk of the bridge collapsing, construction units jacked the girders of Column P6 of the left bridge for three months, and reconstructed the bridge decks of Column P6 of the right bridge. According to the scouring conditions at the bridge site, the circular columns P2, P3, P4, and P5 of the left bridge were adopted as research subjects. Column P2, located on a highland river bank, was employed as the foundation specimen fully embedded in soil, and no scouring was observed around the caisson. Column P3 was the benchmark specimen, and the caisson foundation possessed a 1.2 m exposure length. Column P4, which represents the scoured specimen, had a caisson foundation with a 2.8 m exposure length. To clearly reflect the effects of scouring on the seismic performance of the columns as well as to ensure the safety of the experiment, a further depth of 1.2 m was uncovered, giving a total exposure length of 4 m (approximately one third of the caisson

2 Description of bridge column and caisson foudation specimen The Niu-Dou Bridge was located in Yilan County, Taiwan on the No. 7C Provincial Route. The bridge was adopted for the experiments before demolished in December 2010. The Niu-Dou Bridge consisted of two independent bridge structures, each supporting different traffic flow as shown in Fig. 1. Both bridges had seven spans with pre-stressed concrete I-girder-type decks. The left bridge, which was built in 1995, has a total length of approximately 256.2 m with a net width of 5 m (total width = 5.7 m). The substructure consists of a circular column with a diameter of 1.8 m. The circular caisson

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Fig. 1 Niu-Dou Bridge

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Kuo-Chun Chang et al.: Seismic performance of an existing bridge with scoured caisson foundation

length). Column P5 was selected to study the behavior of a caisson foundation embedded in gravelly soil. The caisson foundation had 1 m exposure. After being hit by many typhoons in the past, the original 4 m diameter circular section was enlarged to 5 m from the caisson top to a depth of 2.8 m to prevent flood scouring. According to the on-site measurement results, the total height of the Columns P2, P3, and P4 were 9.54 m, 10.59 m, and 10.37 m, respectively. According to the design drawings, all of the column sections had a diameter of 1.8 m with the concrete cover thickness of 0.06 m. Thirty D32 longitudinal steel bars were used, and had a 3 m lapsplices in the middle part of the columns. The D16 steel bar was used as lateral reinforcement. The spacing of the lateral reinforcement was 0.2 m at the lower part of the column for 3.6 m length. The same spacing was applied at the upper part of the column as well. In the middle part of the column, the spacing of the lateral reinforcement was 0.3 m for 2.4 m length. The lateral reinforcement was overlapped along 0.6 m without seismic hooks or anchorage to the core. The material properties of the concrete and steel bars, as shown in Table 1, were obtained using sampling measurements at the end of the experimental tests. Geometric and material properties of Column P5 were not measured, hence it is assumed to be the same as the properties of Column P3.

3 Experimental program 3.1 On-site measurement On-site measurement included geotechnical prospecting and the acquisition of the 3D coordinates of the bridge. The full-sampling geotechnical testing was conducted at the bridge site (Chiou et al., 2012). The test results showed that soft gravel, which comprised a mixture of brown-gray to gray gravel of coarse and fine textures with sand silts or silt-like sands, was present 20 m below the ground surface. The N-values of the soft gravel were calculated using the Standard Penetration Test and were found to be generally higher than 50. Furthermore, the shear wave velocity profile obtained through the down-hole borehole seismic measurement was 247 m/s (depth: 0−3 m) and 494 m/s (depth: 3−19.5 m). These results indicate that the geological condition of the soil surface was extremely dense. The 3D coordinates of the bridge were acquired using LiDAR technology. The measurement range and accuracy of the laser scanner was 2000 m and 5 mm, respectively. The original 3D point cloud data was Table 1

processed using coordinate conversion, image overlay, and color integration to form a 3D diagram (point cloud) of the bridge. The plan and elevation views of the projection were adopted as the basis for designing the loading frame and reaction frame. 3.2 Processing of the girder and the expansion joint interface The superstructure was not removed to preserve the actual axial load on the columns. However, the interface due to continuity of bridge decks and safety guards as well as concrete shear keys must be appropriately removed to ensure that the columns are not affected by the additional, lateral stiffness and strength of the neighboring girders and expansion joints. 3.3 Design and analysis of the supplemental frames Based on on-site measurement, the elevation intervals for each cap beam possessed sufficient space for actuator allocation. However, the offset between the left and right columns must be accommodated by installing a U-shaped loading frame, as shown on the cap beam for both the pier wall and the circular column in Fig. 2. Specifically, finite element analysis (FEA) was conducted to examine their stress conditions. The results showed that under the effect of simulative loading, stress at any point of the loading frame produced values that were all lower than half of the yield stress of the steel material. The column wall was employed as a reaction wall because its lateral stiffness was significantly larger than that of the circular column. Considering that the column wall was constructed over 50 years, an A-shaped reaction frame was added to avoid any unexpected damage during the experiments. The base of the frame was fixed on the extended caisson foundation, and the top part of the A-frame was connected in parallel to the U-frame; together, the reaction force that the target column required during pushover could be generated. If the maximum lateral strain was assumed to be 1000 kN, stress at any point of the loading frame produced values that were all lower than half the yield stress of the steel material. Furthermore, the ratio of the initial stiffness for the column wall and the reaction frame stiffness was 1:3.66. In other words, during the initial stages of the experiment, the column wall primarily withstood the reaction force, and as the stiffness of the column wall decreased, the contribution made by the A-frame correspondingly increased. Moreover, the M24 chemical anchor bolt and foundation extension was adopted as the

Measured material properties

Material

153

P2

P3

P4

Concrete

fc' (MPa)

33.4

36.8

31.4

Longitudinal reinforcement

fy (MPa)

365.4

471.8

437.0

Lateral reinforcement

fyt (MPa)

341.0

424.5

354.0

154


Fig. 2 Test setup for Column P3

connector for the A-frame base. The FEA results showed that the tensile stress of the outer chemical anchor bolt was within the stress range permissible for a single chemical anchor bolt.

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superstructure, comprising the combined weight of the pre-stressed girder, concrete bridge decks, asphalt pavement, and the concrete safety guard, was applied. The design diagram indicated an estimated force of 2855 kN. The lateral force was exerted by two actuators with a combined capacity of 2000 kN. As a control, the loading protocol was applied to one actuator, with the displacement calculated using the temposonic sensor. Figure 3 showed the test setup on the caisson foundation of Column P5 (Chiou et al., 2012). The caisson foundation was uncovered to 1 m depth for the installment of the loading system. Two hydraulic jacks in series were used to apply the axial load on the caisson foundation. LVDTs were also used to measure relative displacements on caisson foundation on the left bridge and the caisson foundation of the right bridge, which acted as the reaction wall. A load cell was placed in line with the hydraulic jacks to measure the total lateral load applied. 3.5 Loading procedure This study employed a cyclic loading experiment as the primary test for Columns P3 and P4, and the loading protocol was conducted at drift ratios of 0.25%, 0.375%, 0.5%, 0.75%, 1.0%, 1.5%, 2.0%, and 3.0%, as shown in Table 2. Two cycles were performed at each drift angle. Considering the limitations to the stroke of the actuator, cyclic loading was pulled to 3% drift before it was further pushed to a drift of 5%. Prior to the single-cycle pushover test, a pseudo-dynamic test was conducted on Column P2. An artificial time-history compatible with the design response spectrum according to the seismic design specification for highway bridges (MOTC, 2009) was developed. Because the Niu-Dou Bridge was situated in Yilan County, the data pertaining to the earthquake event on March 31, 2002 (Richter scale of 6.8) in the Niu-Dou region of Sansing Township were used in this study. The design parameters were SsD = 0.8 and S1D = 0.45. The ground motion record was selected from station ILA025. In the experiment, the original time history for east-west (PGA = 90.74 Gal) and southnorth (PGA = 118.62 Gal) directions was assembled into a new ground acceleration. Subsequently, an artificial time history compatible with the design response spectrum was created. Considering the total duration of the original record was relatively long, and the time Side view

Tilt meter P5FL

Steel plate

31.5 cm 35 cm

30 cm

LVDT

LVDT Steel plate

Reference beam 4.9 MN hydraulie jack

Steel box 35 cm

LVDT

Load cell

The measurement system comprised a temposonic sensor, string potentiometer, linear variable differential transformer (LVDT) and tilt meter. The test setup is presented in Fig. 2 using Column P3 as an example. Because the part of the column wall that exerts the reaction force also produces lateral deformation, a reference frame was added downstream of the columns, and the reference frame and the temposonic sensor were adopted as the basis for measuring the absolute displacement, to ensure that the columns would be able to regulate the displacement load during pushover. The string potentiometer measured the displacement of the caisson foundation according to the depth at which the foundation was exposed. It was installed at an elevation level specific to the caisson foundation of the circularcolumn and the column wall and connected to the reference frame. To understand the connection between the chemical anchor bolt, the reaction frame, and the column wall foundation, an LVDT was placed at the base of the reaction frame to measure the displacement of the interface, which was then employed to monitor the safety of the reaction force during the experiment. An LVDT was also installed at the connection point of the loading frame and the two column cap beams to determine the displacement of the interface. Moreover, by using the proportion of the temposonic sensor and the actuator extension, the amount of deformation in each column wall could be calculated. A tilt meter was placed at the base of the column, at the top of the foundation of the two columns, the top part of the column, and on the reference frame to obtain the following information: the curvature distribution of the plastic hinge region on the target column, the angle of twist of the foundation of the two column columns and top part of the column, and the rotation of the reference frame. In this experiment, vertical force from the

1m

3.4 Measurement system and load system

4.9 MN hydraulie jack

River bed

Fig. 3 Test setup for Column P5 (Chiou et al., 2012)

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155

Table 2 Loading protocol Column

Item

P2

P3,P4

Target drift ratio (+/–%) 0.25

0.375

0.5

0.75

1.0

1.5

2.0

3.0

5.0

Displ.(cm)

-

-

-

-

-

-

-

-

51.50

Cycles

-

-

-

-

-

-

-

-

1

Displ. (cm)

2.58

3.86

5.15

7.73

10.30

15.45

20.60

30.90

51.50

Cycles

2

2

2

2

2

2

2

2

1

limitation of the on-site operation and system stability, only the main 40 s vibration record was adopted as the input ground acceleration for the experiment. The input ground motion for pseudo-dynamic test is shown in Fig. 4. The loading protocol for the lateral load test on the caisson foundation of Column P5 was load controlled, which consists of five loading-unloading cycles. The applied load were 638, 1285, 3296, 3590, and 4220 kN.

4 Experimental results and discussion The pseudo-dynamic test was first conducted on Column P2. Figure 5 presents the displacement history with the maximum positive and negative displacements of 0.09 m and 0.07 m, respectively. The hysteretic loop is shown in Fig. 6 (a). Overall, the specimens demonstrated elastic responses, although slight nonlinearity was observed at maximum displacement. This result indicates that Column P2 is secure, even when the simulated earthquake was set to a peak ground acceleration of 0.32 g. Figure 7 presents the cracking patterns of the columns generated after the pseudo-dynamic test. Eight major cracks were presented on the upstream side, and 10 cracks were observed on the downstream side; however, the crack widths were minimal. Following the pseudodynamic test, a one-cycle pushover test was conducted. The lateral strength and displacement results are shown in Fig. 6(b), and the damage is displayed in Fig. 8. The concrete cover and lateral stirrups on both the top and bottom regions exhibited substantial spalling. The damaged area was approximately up to a height of 1 m from the base, and the main steel bar at the base of the

column demonstrated significant buckling. The lateral strength and displacement of Column P3 under cyclic loading is presented in Fig. 9. Figure 10 presents the deformation and the crack patterns of the columns at maximum displacement during pushover. In addition, major concrete spalling was observed on the downstream side. The damaged area was approximately up to a height of 1.1 m from the base, and the main steel bar of the column base demonstrated significant buckling. On the upstream side, no concrete spalling was identified, but several cracks, approximately 0.2 m apart and 6 mm wide, were observed on the surface and parallel to the stirrup. The highest point at which the major crack was located was approximately 1.6 m from the base. After cover concrete were removed, the thickness of the cover concrete was recorded at 0.1 m, which preliminarily explains why the steel bar of that region did not buckle. Figure 11 shows the hysteretic loop of Column P4 under cyclic loading, and Fig. 12 presents the deformation and crack patterns of the columns at maximum displacement during pushover. The experimental procedure for Column P4 was similar to that for P3. The concrete protective cover and the lateral stirrups on both the upstream and downstream regions exhibited substantial spalling. The damaged region was approximately up to 1.1 m from the base and the main steel bar of the column base demonstrated significant buckling. When the right bridge was constructed, no regulations regarding seismic design were specified. Furthermore, when comparing the bridge to the original design, the lateral steel bar had a 0.6 m long lap joint, but this was not connected to a seismic hook. Consequently, the desired confinement effect was not achieved from 100 Displacement (mm)

Acceleration (g)

0.4 0.2 0 -0.2 -0.4

0

5

10

15

20 25 Time (s)

30

Fig. 4 ILA025 earthquake ground motion

35

40

20 0 -50 -100

0

5

10

15

20 25 Time (s)

30

35

40

Fig. 5 Displacement time history for Column P2 from pseudo-dynamic test

156


-6

-2

Drift ratio (%) 0 2

4

6

-6 1.0

P2 column

Lateral force (MN)

Lateral force (MN)

1.0

-4

0.5

0

Pull

Push

0

-0.5

-1.0

-1.0 -400

-200 0 200 400 Lateral displacement (mm) (a) Column P2 from pseudo-dynamic test

-2

Drift ratio (%) 0 2

600

Pull

-600

Push

-400

-200 0 200 400 Lateral displacement (mm) (b) Column P2 from cyclic loading test

(b) Downstream side

(c) Upstream side

Fig. 7 Damage of Column P2 at bottom from pseudo-dynamic test

(a) Side view

6

P2 column

Fig. 6 Hysteresis diagram for Column P2

(a) Side view

4

0.5

-0.5

-600

-4

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(b) Downstream side

(c) Upstream side

Fig. 8 Damage of Column P2 at bottom from cyclic loading test

600

Suppl.1


-6

-4

Drift ratio (%) 0 2

4

-6

6

0.5

0

Pull

Push

0

-0.5

-1.0

-1.0 -400

-200 0 200 400 Lateral displacement (mm)

-2

Drift ratio (%) 0 2

4

6

P4 column

0.5

-0.5

-600

-4

1.0

P3 column

Lateral force (MN)

Lateral force (MN)

1.0

-2

157

Pull

-600

600

Push

-400

-200 0 200 400 Lateral displacement (mm)

600



the steel bar embedded in the core of the concrete. Therefore, lateral stirrup spalling was observed. Overall, comparing the lateral strength and displacement between the three columns, the resulting damage was similar, despite the differences in boundary conditions. Furthermore, the maximum lateral strength of P2 (1069 kN), P3 (1058 kN), and P4 (1125 kN) did not differ significantly, suggesting that the construction quality for the bridge structure was reliable. Column P2 was slightly damaged because of the pseudo-dynamic test, although the maximum strength and deformation was not affected. When the lateral strength of Column P3 was reduced, a platform section was developed. In addition, for every displacement, the strength and stiffness at the second cycle was similar to that at the first cycle. This result indicates stable energy dissipation. Finally, the presence of the gravel and caisson enabled

Column P4 to produce a substantially large lateral stiffness. Although a 4 m deep foundation was exposed, the structural behavior was similar to that of the fullyembedded specimen. Figure 13 showed the pushover result of a caisson foundation on Column P5. The curve was considered very stiff given a very small displacement of 0.0155 m at maximum load of 4220 kN. The stiff behavior might come from the large lateral resistance provided by the dense gravelly soil. Moreover, the aggregate interlocking of the gravel soil particle would increase the friction on particles as well as friction of soil-caisson interface. Although the caisson has a very large stiffness, the behavior was highly nonlinear within small displacement, particularly during the unloading and reloading stage.

(a) Side view

(b) Downstream side

Fig. 10 Damage of Column P3 at bottom from cyclical loading test

(c) Upstream side

158


(a) Side view

(b) Downstream side

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(c) Upstream side

Fig. 12 Damage of Column P4 at bottom from cyclical loading test

5

Force (MN)

4 3 2 1 0 0

Fig. 13

5

4

8 12 Displacement (mm)

16

Load deflection curve of caisson foundation of Column P5 (Chiou et al., 2012)

Finite element simulation

The analytical model was constructed in the OpenSees platform (2005) to simulate the experimental result. The complete highly nonlinear analytical model, as shown in Fig. 14, consists of nonlinear fiber elements to represent the nonlinear flexural behavior of the column, springs in series at the column base to simulate bond-slip and strength degradation due to buckling of longitudinal reinforcement, elastic elements to represents massive caisson foundation, and nonlinear winkler springs to represent the soil behavior. 5.1 Force based fiber beam column element The flexural behavior of the column was modeled using force-based fiber beam-column elements. This element was able to model the spread of plasticity along the members. The constitutive stress-strain relationship of concrete and steel that was assigned into fiber sections is shown in Fig. 15. The concrete fibers were assigned

using the constitutive stress-strain relationships proposed by Mander et al. (1988), which can be represented using the “Concrete 04” material model in OpenSees. The ultimate strain of unconfined concrete was assumed to be 0.004. However, the ultimate strain for the confined concrete was assumed to be infinite in order to eliminate the localization phenomenon, which occurs frequently in the force based beam-column element. The steel fibers were assigned using a bilinear constitutive stress-strain relationship. The “Hysteretic” material model in OpenSees was chosen to represent the bilinear steel model with a strain-hardening ratio equal to 1.5%. Parameter Pinchx and Pinchy in the steel model were set to 0.4 and 0.6, respectively, in order to accurately model the global pinching behavior. 5.2 Rotational slip spring element An elastic rotational slip spring was introduced at the end of the fiber beam column element to account for the additional rigid body rotation caused by strain penetration or bond-slip of the longitudinal reinforcing bars at the column foundation interface. The elastic

Fiber beam column element Elastic beam column element

Rotational slip spring element Shear spring (Limit state material with bucking limit curve) Rigid link Soil spring

Fig. 14 Analytical model


30

30

Stress (MPa)

45

15

159

800

45

Stress (MPa)

Stress (MPa)

Suppl.1

15

400 0 -400

0

0

2

4 Strain (10 ) (a) Unconfined concrete

6

0

0

5

-3

10 Strain (10-3)

-800 -100

15

-50

(b) Confined concrete

0 50 Strain (10-3) (c) Bilinear steel

100

Fig. 15 Constitutive stress strain relationship

rotational stiffness proposed by Elwood and Eberhard (2009) was selected to be incorporated on the analytical model. The equation of the rotational stiffness was as follows: K slip

8u M y  d b f y y

(1)

where db is the nominal diameter of the longitudinal reinforcement, fy is the yield stress of longitudinal reinforcement, My is the effective yield moment, φ y is the effective yield curvature, and u is the uniform bond stress; which can be approximated using 0.8 f c' . 5.3 Buckling spring element The buckling spring was used to represent the behavior of the reinforced concrete bridge pier after the bar buckling failure was detected. In this study, the available shear spring with shear limit state developed by Elwood and Moehle (2003) was used as the buckling spring model. The hysteretic parameters on the shear spring such as: pinchx, pinchy, damage 1, damage 2, and beta, were set as 1.0, 1.0, 0.0, 0.0, and 0.4, respectively. In order to activate the shear spring at the bar buckling failure point, the original shear limit curve, developed by Elwood and Moehle (2003). needed to be shifted to the buckling point proposed by Berry and

Shear spring response V

Eberhard (2005), as shown in Fig. 16, using the delta function. The buckling limit equation is shown in Eq.(2) . d   bb P  L   (2)  %   3.25 1  ke_bb eff b  1  '  1  L D   Ag f c   10 D  

where ∆bb/L is the drift ratio at buckling failure, Ke_bb = 40 for rectangular-reinforced columns and 150 for spiralreinforced columns, but should be taken as 0 for columns in which s/db exceeds 6, ρeff = ρsfyt/f 'c, ρs is the volumetric transverse reinforcement ratio, fyt is the yield stress of the transverse reinforcement, L is the distance from the column base to the point of contra flexure, and D is the column depth. After the onset of bar buckling, the shear strength of the pier was assumed to decrease until all shear strength capacity was lost. The collapse displacement, ∆c, is assumed as twice ∆bb. Hence, the degrading slope of the total response, Ktdeg, can be estimated as follows: t K deg 

Vu

  c   bb 



Vu  bb

(3)

The degrading slope in shear spring response, Kdeg, was empirically taken as 0.5 Ktdeg.

Beam - column + bond slip response

Total response shear limit curve

Pre-failure backbone

shifted shear limit curve

Vu Kdeg

Ktdeg

Kunload

Δs

Δf+bs

Fig. 16 Shear spring with shifted shear limit curve

Δs

Δbb

Δc

Δ

160


5.4 Elastic caisson element and soil spring element Considering a very large geometry of the caisson foundation in comparison with the bridge column, it is assumed that the caisson foundation will behave elastically during the experimental test; and hence, the elastic beam column element was used to represent the behavior of the foundation. Since the caisson possessed a large diameter, frictional forces between the foundation and soil, which occur at the foundation faces, might provide additional moment resistance. Therefore, rigid link elements with a length equal to the radius of the caisson were introduced to connect the elastic caisson element with the soil springs as shown in Fig. 17. The soil reaction was modeled using a Winkler spring model. Figure 17 shows the arrangement of the soil spring elements on the caisson foundation. In general, there are four types of soil springs to be used in the analytical model: the spring which represented horizontal subgrade reaction of soil, the spring which modeled the friction around the caisson surface, the spring which modeled the base shear, and the spring which provided vertical end bearing to the foundation. The load deformation response for horizontal subgrade reaction, frictional shear and base shear were determined from experimental results of the plate loading test and direct shear test with a few modifications of the stiffness and strength. The springs' properties were directly adopted from Chiou et al. (2012). On the other hand, the end bearing was calculated based on the equation introduced by Meyerhof (1976).

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Uniaxial material “PySimple1” was placed in a horizontal direction to provide subgrade reaction of soil (Py spring 301, 302, 303,…; Py spring 501, 502, 503,…) . The uniaxial material “TzSimple1” was used to represent the vertical frictional forces on the front face (Tz spring 401, 402, 403,…), vertical friction on the back face (Tz spring 601, 602, 603,…), vertical friction on the side faces (Tz spring 2, 3, 4,…), horizontal friction on the side faces (Tz spring 102, 103, 104,…), as well as base shear force of the caisson foundation (Tz spring 101). Material “QzSimple1” was used to provide the axial end bearing (Qz spring 1). Figure 18 shows the total force and displacement of the Py spring. 5.5 Loading condition For all specimens, the axial load was calculated based on the tributary area of the superstructure. The selfweight of the substructure and the caisson foundation was also taken into account and lumped on several points on the beam column elements representing the distributed load. After the axial loads were finished, the next loading was applied. For Column P2, the time history analysis was conducted. The mass source was assumed to be provided by the superstructure only, which is 285.5 kg. The damping ratio was assumed to be 5%. After the time history analysis completed, the cyclic loading analysis was simultaneously conducted. For Columns P3, P4 and P5, after the completion of the axial loading analysis, the cyclic loading analysis were conducted with the loading protocols set as the same target displacement as achieved in the experimental test.

6 Analytical results

Tz spring 455

Tz spring 655 Py spring 555

Tz spring 55 Tz spring 555

Py spring 355




The pushover result on the caisson foundation of Column P5 is shown in Fig. 19. A comparison with the experimental result showed that the analytical result can produce an acceptable envelope curve. However, there are still some discrepancies between the test result and

3.5 Tz spring 653 Py spring 553



3.0



Py spring 302 Tz spring 401


Tz spring 402

Tz spring 101 Qz spring 1

Fig. 17 Soil spring elements

P (MN/m)

2.5 2.0 1.5 1.0 0.5

Depth = 1 m

Py spring 301 0

0

0.05

0.10

Depth = 2 m 0.15 y (m)

0.20

Fig. 18 Soil springs

Depth = 4 m 0.25

0.30

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analytical result, such as: the ultimate force was over predicted by 10%, the predicted unloading trend did not perfectly follow the test result although the residual displacements after the unloading were close enough, and the reloading forces were over-predicted. Given that the ultimate forces of the columns were up to approximately 1000 kN and the analytical result produced a very close envelope up to 3500 kN force, which corresponds to 10 mm displacement, the soil model was considered good enough to be used in the analysis of the rest of the column specimens. The displacement history of Column P2 is presented in Fig. 20. It is observed that in the time range from 20 s to 30 s, the analytical model could not capture the amplitude and phase angle very well. The inconsistency of the phase and amplitude in this time range might come from the disagreement between the unloading and reloading part of the soil properties. Despite the inconsistency in that 10 s range, the analytical model could predict the displacement history very well. The maximum and

161

minimum displacement of the experimental test data was observed to be 90.6 mm at 15.54 s and -69.94 mm at 15.16 s, respectively, while the analytical model predicted the maximum and minimum displacement of 89.68 at 15.56 s and -67.884 at 15.16 s. Slight variations are considered acceptable in engineering practice. Moreover, the residual displacement after 30 s is well predicted. The overall hysteresis loop is presented in Fig. 21(a). The hysteresis loop result of the analytical model also closely followed the test result. Due to the difficulties in predicting the actual behavior of soil, this result is considered acceptable. Figure 21(b) show the one large cycle push pull analysis result after the time history analysis are completed. The result showed that the analytical model could predict the test data accurately in terms of initial stiffness as well as ultimate force. The analytical model also predicted that buckling will occur at 4.76 % drift; and hence, the strength will drop afterward. This model clearly described the post

5

Force (MN)

4

Displacement (mm)

Test result Analytical result

3 2

100 50 0 Test result

-50

Analytical result

-100 0

1

5

10

15

20 Time (s)

25

30

35

40

Fig. 20 Displacement history for Column P2 0

0

4

8 Displacement (mm)

12

16

Fig. 19 Pushover results of caisson foundation Column P5

1.2

1.2 Test result

Test result Analytical result

0.6 Force (MN)

Force (MN)

0.6

0

0

-0.6

-0.6

-1.2 -120

Analytical result

-60

0 60 Displacement (mm) (a) Time history analysis

120

-1.2 -600

Fig. 21 Hysteresis Loop of Column P2

-200 0 200 Displacement (mm) (b) Pushover analysis

600

162


buckling failure behavior very well. The arrangement of the sensors during the large cycle push experimental test on Column P2 is shown in Fig. 22(a). LVDT 49 and 50 were connected to the caisson foundation of the bridge with the reaction system. Therefore, only LVDT line 49 and line 50 provided a stable symmetric loop while line 47 and 48 had an asymmetric loop due to instability of the reference frame. Figure 22(b) shows the comparison of the analytical result with the test result. The analysis result shows that the maximum displacement predicted on the caisson foundation on ultimate force is very close to the experimental result. The stiffness is also close enough. The overall result is considered acceptable. The analytical result of the benchmark Column P3 is shown in Fig. 23(a). The overall hysteretic behavior predicted by the analytical model was accurate enough although there are microscopic differences on the initial stiffness. The ultimate strength, unloading forces, reloading forces, and pinching behavior are well predicted. The simulation predicted the strength drop due to buckling would occur at 4.94 % drift. Since the actual experimental test was conducted until 5% drift, the strength drop was not clearly shown on the analytical result. The analytical result of Column P4 with a 4 m exposed caisson foundation is shown in Fig. 23 (b). The same good result was also achieved in predicting the behavior of this particular specimen. The ultimate strength, unloading forces, reloading forces, and pinching behavior are well predicted. Although there is a slight difference on the initial stiffness, the overall hysteresis loop is well predicted. The analytical model predicted the buckling point would occur at 4.87% drift. However, the test was terminated at drift 4.6%. Therefore, the

17,18 LVDT (100 °C) 9, 10

11,12 LVDT (100 °C)

13,14 LVDT (100 °C)

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strength degradation due to buckling behavior was not simulated on this specimen.

7 Parametric study of Niu-dou Bridge The analytical model was verified and proved to be accurate enough to predict the behavior of the insitu tested bridge column. This section provides a parametric study to assess the scouring effect of the caisson foundation on the seismic performance of the bridge column. Column P2 was selected as the subject considering that the analytical model could predict the displacement history as well as the hysteresis loop well enough. The artificial generated earthquake ground motion ILA025, which was used as the input ground motion in the pseudo-dynamic test, was used for the parametric study. In this parametric study, the scouring depth was changed from full embedded, scoured to 2 m deep, 4 m deep, 6 m deep, 8 m deep, and 10 m deep. The initial period of the structure system was calculated to be 0.56 s, 0.58 s, 0.62 s, 0.66 s, 0.75 s, and 1.2 s with respect to the exposure length. For each case, incremental dynamic analysis was conducted using both earthquakes with PGA scaled from 0.1 g to 1 g. The results of incremental dynamic analysis (IDA) using the ILA025 artificial earthquake ground motion was shown in Fig. 24. It is clearly seen that when the column foundation is scoured to 2 m deep, the column and foundation still behave like the one without the scouring condition. When the scouring reaches 4 m deep, the additional flexibility introduced by the exposed caisson foundation contribute to the reduction of the shear force on the column. As the result, the column will reach yielding and maximum force at a later stage. When

19,20

1.5

LVDT (100 °C) 15,16

1.5 Test result

1.0

Test result

1.0

Analytical result

7, 8 LVDT (100 °C) 41, 42

49, 50 LVDT

(a) Sensor layout

Fig. 22

28 27 26 25 24 23 22 21

36 35 34 33 32 31 30 29

47

0.5 Force (MN)

Force (MN)

0.5

Analytical result

0 -0.5

0 -0.5

-1.0

-1.0

-1.5 -5.0 -2.5 0 2.5 5.0 Displacement (mm)

-1.5 -5.0 -2.5 0 2.5 5.0 Displacement (mm)

(b) Line 49

Displacement on caisson foundation of Column P2 during cyclic loading

(c) Line 50

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1.2

1.2 Test result

Test result

Analytical result

0.6 Force (MN)

Force (MN)

0.6

163

0

-0.6

Analytical result

0

-0.6

-1.2 -600

-200 0 200 Displacement (mm) (a) P3 Column

-1.2 -600

600

-200 0 200 Displacement (mm) (b) P4 Column

600

Fig. 23 Hysteresis Loop of Column P3 and P4

1.2

1.0

0.8

PGA (g)

Force (MN)

0.9

0.6 Exposed length 0.3

0

0.2

2m

4m

6m

8m

10 m

0.4

0.6 0.8 PGA (g) (a) Maximum shear force at bridge column

0

1.0

1.0

1.0

0.8

0.8

0.6

0.6

Exposed length

0.4

0.2

0

0

200

0m

2m

4m

6m

8m

10 m

400 600 800 Displacement (mm) (c) Maximum displacement at caisson tip

Exposed length

0.4

0.2

PGA (g)

PGA (g)

0

0m

0.6

0

4m

6m

8m

10 m

1000 1500 Displacement (mm) (b) Maximum displacement at column tip

2000

Exposed length

0.2

0

2m

500

0.4

1000

0m

0

0m

2m

4m

6m

8m

10 m

200

400 600 800 Displacement (mm) (d) Maximum column tip displacement relative to caisson tip

Fig. 24 IDA result of Column P2 with different exposed length of caisson foundation

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the scouring reaches 6 m deep, the shear force only reaches 900 kN at PGA 1 g, which means that the column behaves elastically although the displacement becomes larger. When the caisson foundation is exposed by 8 m, the system becomes highly flexible, and the column tip displacement diplaced more than 1 m at PGA 1 g. At 10 m of scouring, the the analysis was increased to 0.01 g after 0.5 g and terminated at 0.56 g, since the additional increment of the PGA would create a large displacement and cause an error.

8

Conclusions

The seismic performance of three circular piers built in 1995 was studied experimentally based on in-situ tests. It was found that the pier (diameter 1.8 m, aspect ratio 5.72, longitudinal reinforcement ratio 1%, and lateral reinforcement ratio 0.294%) with a fully embedded caisson remained elastic under a 0.32 g excitation from a pseudo-dynamic test, and the deformation capacity could reach a 5% drift ratio, while providing lateral strength of 1000 kN. Special consideration was given to investigate the effects of scouring with an exposed caisson foundation. The experimental test result predicted that when the exposed caisson foundation reached 4 m, the column behavior only has a minor difference when compared to the fully embedded column. An analytical model based on a finite element model was set up to consider nonlinearity on flexural, bond-slip, buckling mechanism, and soil behavior. The analytical model could accurately predict the test results. A parametric study has shown that the scouring effect will have a positive effect on lowering the shear force introduced to the column. However, scouring will have negative effect on increasing the displacement demand on the structure depending on the ground motion characteristics.

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